Innovative Infrastructure Solutions using Geosynthetics: Proceedings of the 3rd GeoMEast International Congress and Exhibition, Egypt 2019 on Sustainable Civil Infrastructures – The Official International Congress of the Soil-Structure Interaction Group in Egypt (SSIGE) [1st ed. 2020] 978-3-030-34241-8, 978-3-030-34242-5

Table of contents :
Front Matter ....Pages i-x
Giant Geotextile Tube Applied to the Temporary Cofferdam Reclamation Construction for a New-Build Container Base at Harbor in Taiwan (Peter Huang, Belinda Lai)....Pages 1-8
Effect of Strain Rate on Cyclic Behavior of Pond Ash Reinforced with Geocell (Swaraj Chowdhury, NiharRanjan Patra)....Pages 9-21
Developments in MSE Wall Research and Design (Richard J. Bathurst)....Pages 22-50
Cemented Lateritic Soil as Base Material Improvement Using Compression (Kennedy Chibuzor Onyelowe, Talal Amhadi, Charles Ezugwu, Eze Onukwugha, Henry Ugwuanyi, Ifeoma Jideofor et al.)....Pages 51-61
Geomembrane Stress Cracking Resistance Depending on the Polymer Used (J. M. Muñoz Gómez)....Pages 62-67
A Geomembrane Liner to Stop Water Seepage in an 8 km Long Embankment: The Bill Young Reservoir Case History (Alberto Scuero, Gabriella Vaschetti, John Wilkes)....Pages 68-81
Experimental Research and Application of Geopolymer in Soft Soil Foundation Treatment (Jialiang Yao, Haojie Qiu, Hua He, Xin Chen, Guiyu Hao)....Pages 82-94
Effect of Offset Distance on Tiered Reinforced Soil Retaining Wall Subjected to Dynamic Excitation (Sudipta Sikha Saikia, Arup Bhattacharjee)....Pages 95-107
Review of Process Control and Assurance for Optimized Seaming Condition Optimization of Woven Geotextiles to Improve Stability in Soft Soil Structure (Han-Yong Jeon)....Pages 108-120
Celebrating Reinforced Soil Structures (Chaido Doulala-Rigby)....Pages 121-150
Back Matter ....Pages 151-151

Citation preview

Sustainable Civil Infrastructures

Fumio Tatsouka Erol Guler Hany Shehata J. P. Giroud Editors

Innovative Infrastructure Solutions using Geosynthetics Proceedings of the 3rd GeoMEast International Congress and Exhibition, Egypt 2019 on Sustainable Civil Infrastructures – The Official International Congress of the Soil-Structure Interaction Group in Egypt (SSIGE)

Sustainable Civil Infrastructures Editor-in-Chief Hany Farouk Shehata, SSIGE, Soil-Interaction Group in Egypt SSIGE, Cairo, Egypt Advisory Editors Khalid M. ElZahaby, Housing and Building National Research Center, Giza, Egypt Dar Hao Chen, Austin, TX, USA

Sustainable Infrastructure impacts our well-being and day-to-day lives. The infrastructures we are building today will shape our lives tomorrow. The complex and diverse nature of the impacts due to weather extremes on transportation and civil infrastructures can be seen in our roadways, bridges, and buildings. Extreme summer temperatures, droughts, flash floods, and rising numbers of freeze-thaw cycles pose challenges for civil infrastructure and can endanger public safety. We constantly hear how civil infrastructures need constant attention, preservation, and upgrading. Such improvements and developments would obviously benefit from our desired book series that provide sustainable engineering materials and designs. The economic impact is huge and much research has been conducted worldwide. The future holds many opportunities, not only for researchers in a given country, but also for the worldwide field engineers who apply and implement these technologies. We believe that no approach can succeed if it does not unite the efforts of various engineering disciplines from all over the world under one umbrella to offer a beacon of modern solutions to the global infrastructure. Experts from the various engineering disciplines around the globe will participate in this series, including: Geotechnical, Geological, Geoscience, Petroleum, Structural, Transportation, Bridge, Infrastructure, Energy, Architectural, Chemical and Materials, and other related Engineering disciplines.

More information about this series at http://www.springer.com/series/15140

Fumio Tatsouka Erol Guler Hany Shehata J. P. Giroud •

•

•

Editors

Innovative Infrastructure Solutions using Geosynthetics Proceedings of the 3rd GeoMEast International Congress and Exhibition, Egypt 2019 on Sustainable Civil Infrastructures – The Official International Congress of the Soil-Structure Interaction Group in Egypt (SSIGE)

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Editors Fumio Tatsouka University of Tokyo Tokyo, Japan Hany Shehata Soil-Structure Interaction Group in Egypt (SSIGE) Cairo, Egypt

Erol Guler Bogazici University Besiktas, Turkey J. P. Giroud US National Academy of Engineering Paris, France

ISSN 2366-3405 ISSN 2366-3413 (electronic) Sustainable Civil Infrastructures ISBN 978-3-030-34241-8 ISBN 978-3-030-34242-5 (eBook) https://doi.org/10.1007/978-3-030-34242-5 © Springer Nature Switzerland AG 2020 This work is subject to copyright. All rights are reserved by the Publisher, whether the whole or part of the material is concerned, specifically the rights of translation, reprinting, reuse of illustrations, recitation, broadcasting, reproduction on microfilms or in any other physical way, and transmission or information storage and retrieval, electronic adaptation, computer software, or by similar or dissimilar methodology now known or hereafter developed. The use of general descriptive names, registered names, trademarks, service marks, etc. in this publication does not imply, even in the absence of a specific statement, that such names are exempt from the relevant protective laws and regulations and therefore free for general use. The publisher, the authors and the editors are safe to assume that the advice and information in this book are believed to be true and accurate at the date of publication. Neither the publisher nor the authors or the editors give a warranty, expressed or implied, with respect to the material contained herein or for any errors or omissions that may have been made. The publisher remains neutral with regard to jurisdictional claims in published maps and institutional affiliations. This Springer imprint is published by the registered company Springer Nature Switzerland AG The registered company address is: Gewerbestrasse 11, 6330 Cham, Switzerland

Contents

Giant Geotextile Tube Applied to the Temporary Cofferdam Reclamation Construction for a New-Build Container Base at Harbor in Taiwan . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Peter Huang and Belinda Lai Effect of Strain Rate on Cyclic Behavior of Pond Ash Reinforced with Geocell . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Swaraj Chowdhury and NiharRanjan Patra Developments in MSE Wall Research and Design . . . . . . . . . . . . . . . . . Richard J. Bathurst Cemented Lateritic Soil as Base Material Improvement Using Compression . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Kennedy Chibuzor Onyelowe, Talal Amhadi, Charles Ezugwu, Eze Onukwugha, Henry Ugwuanyi, Ifeoma Jideofor, Chidozie Ikpa, Uzoma Iro, and Benjamin Ugorji

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Geomembrane Stress Cracking Resistance Depending on the Polymer Used . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . J. M. Muñoz Gómez

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A Geomembrane Liner to Stop Water Seepage in an 8 km Long Embankment: The Bill Young Reservoir Case History . . . . . . . . . . . . . Alberto Scuero, Gabriella Vaschetti, and John Wilkes

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Experimental Research and Application of Geopolymer in Soft Soil Foundation Treatment . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Jialiang Yao, Haojie Qiu, Hua He, Xin Chen, and Guiyu Hao

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Effect of Offset Distance on Tiered Reinforced Soil Retaining Wall Subjected to Dynamic Excitation . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Sudipta Sikha Saikia and Arup Bhattacharjee

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Contents

Review of Process Control and Assurance for Optimized Seaming Condition Optimization of Woven Geotextiles to Improve Stability in Soft Soil Structure . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 108 Han-Yong Jeon Celebrating Reinforced Soil Structures . . . . . . . . . . . . . . . . . . . . . . . . . 121 Chaido Doulala-Rigby Author Index . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 151

About the Editors

Fumio Tatsouka received the degree of PhD in Geotechnical Engineering, University of Tokyo, 1973, and is Professor Emeritus of The University of Tokyo and Tokyo University of Science. He was Vice-President, ISSMGE (2001–2005); President, Japanese Geotechnical Society (2007–2008); Vice-President, Japanese Society for Civil Engineers (2005–2006); and Vice-President and President, International Geosynthetic Society (2002–2006 and 2006–2010). He is specialized in the deformation and strength characteristics of geomaterials, ground improvement, and geosynthetic-reinforced soil structures. He received 1994 IGS Award; 1996–1997 Mercer Lectureship, ISSMGE/IGS; 1996, 2000, and 2003 Hogentogler Award, ASTM; 1997 Best Paper Award, the Ground Improvement Journal, ISSMGE: 2007, 2008, 2010, and 2011 Best Paper Award, Geosynthetics International Journal; 2008 Best Paper Award, Geotextiles and Geomembranes Journal; 2011 Bishop Lectureship, ISSMGE, and many others. He wrote a number of technical papers in international journals and did a number of invited lectures at international conferences and symposia of geotechnical engineering and geosynthetic engineering.

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About the Editors

Dr. Erol Guler is Full Professor of geotechnical engineering at Bogazici University, Istanbul, Turkey, since 1989. He acted as Director of Environmental Sciences Institute of Bogazici University between 1996 and 1999 and as Chairman of the Civil Engineering Department between 2004 and 2010. He was Visiting Fulbright Professor at the University of Maryland between 1989 and 1991. He is Leading Geosynthetic Scientist in Turkey, having been IGS Member since 1989. He founded the IGS Turkish Chapter in 2001 and served as its President until 2005 and was reelected as President again in 2011. He was Organizer for the first two national geosynthetic conferences in 2004 and 2006 and is currently Chairman of the 7th congress which will be held in 2017. He was also of Chairman of the 2016 European Regional Conference of IGS, EuroGeo6. He has been Member of the International Standards Organization (ISO) Technical Committee on geosynthetics as Representative of the Turkish Standards Institute since 2002. He is currently Convener of the WG2 of ISO/TC221 (Technical Committee on geosynthetics) and is also Convener of the WG2 of CEN-TC189 (European Committee for Standardization’s Technical Committee on geosynthetics). He is currently International Member of the USA TRB Committee on Geosynthetics. His research has focused mainly on geosynthetic-reinforced walls, and specifically, he conducted research on the use of marginal soils in such structures and their behavior under earthquake loading conditions. His research work includes numerical studies as well as shaking table tests and full-scale tests. He has over one hundred scientific publications. In addition to his research work, he has extensive practical experience, including design work for numerous projects where geosynthetics were used as reinforcement or liners.

About the Editors

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Hany Shehata is Founder and CEO of the Soil-Structure Interaction Group in Egypt “SSIGE.” He is Partner and Vice-President of EHE Consulting Group in the Middle East and Managing Editor of the “Innovative Infrastructure Solutions” journal, published by Springer. He worked in the field of civil engineering early, while studying, with Bechtel Egypt Contracting & PM Company, LLC. His professional experience includes working in culverts, small tunnels, pipe installation, earth reinforcement, soil stabilization, and small bridges. He also has been involved in teaching, research, and consulting. His areas of specialization include static and dynamic soil–structure interactions involving buildings, roads, water structures, retaining walls, earth reinforcement, and bridges, as well as, different disciplines of project management and contract administration. He is Author of an Arabic practical book titled “Practical Solutions for Different Geotechnical Works: The Practical Engineers’ Guidelines.” He is currently working on a new book titled “Soil-Foundation-Superstructure Interaction: Structural Integration.” He is Contributor of more than 50 publications in national and international conferences and journals. He served as Co-chair of the GeoChina 2016 International Conference in Shandong, China. He serves also as Co-chair and Secretary General of the GeoMEast 2017 International Conference in Sharm El-Sheikh, Egypt. 2016 Outstanding reviewer of the ASCE as selected by the Editorial Board of International Journal of Geomechanics.

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About the Editors

Dr. J. P. Giroud Member of the US National Academy of Engineering Dr. J. P. Giroud, Chevalier in the Order of the Légion d’Honneur and Former Professor of geotechnical engineering, is Consulting Engineer, Member of the US National Academy of Engineering, Doctor Honoris Causa of the Technical University of Bucharest, Past President of the International Geosynthetics Society (IGS), Chairman Emeritus and Founder of Geosyntec Consultants, and Chairman of the Editorial Board of Geosynthetics International. He has authored over 400 publications. He coined the terms “geotextile” and “geomembrane” in 1977. He has developed many of the design methods used in geosynthetics engineering and has originated a number of geosynthetics applications. In 1994, the IGS named its highest award “The Giroud Lecture,” “in recognition of the invaluable contributions of Dr. J. P. Giroud to the technical advancement of the geosynthetics discipline”; a Giroud Lecture is presented at the opening of each International Conference on Geosynthetics. In 2002, he became Honorary Member of the IGS with the citation “Dr. Giroud is truly the father of the International Geosynthetics Society and the geosynthetics discipline.” In 2005, he has been awarded the status of “hero” of the Geo-Institute of the American Society of Civil Engineers (ASCE) and has delivered the prestigious Vienna Terzaghi Lecture. In 2005–2006, he presented the Mercer Lectures, a prestigious lecture series endorsed jointly by the IGS and the International Society for Soil Mechanics and Geotechnical Engineering (ISSMGE). In 2008, he delivered the prestigious Terzaghi Lecture of the ASCE. In 2016, he delivered the prestigious Victor de Mello Lecture of the ISSMGE and, in 2017, the prestigious Széchy Lecture, in Budapest. He has 56 years of experience in geotechnical engineering, including 48 years on geosynthetics.

Giant Geotextile Tube Applied to the Temporary Cofferdam Reclamation Construction for a New-Build Container Base at Harbor in Taiwan Peter Huang(&) and Belinda Lai ACE Geosynthetics, Taichung City, Taiwan {peter.huang,belinda.lai}@geoace.com

Abstract. Kaohsiung Harbor is the largest international commercial port in southern Taiwan and also the largest container exporting centre in Taiwan. In response to the growing trend of large-scale container ships in the international shipping and to strengthen the competitiveness of the Kaohsiung Container Hub port in the Asia-Pacific, a new container base in the hinterland of the southern tip of Kaohsiung Harbor is built. As this new land-reclaimed container base will be completed in phases over a period of years, a temporary cofferdam is needed after the 1st phase of reclaimed work (the 1st phase of the reclaimed area is 43 ha and the total area is 162 ha). Because of the deep water (4–5 m deep after dredging), it is recommended to enlarge the size of single geotextile tube to assure the stability of cofferdam and to achieve the goal of reducing the usage of geotextile tubes. The final proposal is to fill hydraulically the giant geotextile tube (here used is ACETube®) with special safeguard of it during the filling process, and stacking in a 2+1 pyramid shape to form a designed cofferdam dimension of more than 7.5 m high, 32 m wide base and total structure length of 700 m. Before filling, the maximum circumference of each ACETube® is 36 m with a length of 34.5 m, and the dimensions of a single ACETube® after filling can reach 4 m high and 16 m wide. In this project, it successfully overcomes the construction difficulty of the 4 m underwater filling procedures of the giant geotextile tube. The use of this giant geotextile tube underwater prevents the cofferdam erosion caused by wave and tide and enhances the stability of the cofferdam to resist the heavy duty work at the backfill area.

1 Introduction 1.1

Project Description

Kaohsiung Harbour is the TOP 15 container harbour in the world, and it handles over 10 millions of twenty-foot equivalent units (TEU) of cargo annually. Therefore, it is important to ensure that the spaces for berth and storage are sufficient for its heavy loading. This case is the new land reclamation project named “Kaohsiung Port Intercontinental Container Terminal Phase II Project (Fig. 1). The existing caisson is outstretched (Fig. 2) and the sea sand is backfilled to form new land (Fig. 3). Then, ACETube® is adopted in the L-shape temporary cofferdam to stabilize the coast and © Springer Nature Switzerland AG 2020 F. Tatsouka et al. (Eds.): GeoMEast 2019, SUCI, pp. 1–8, 2020. https://doi.org/10.1007/978-3-030-34242-5_1

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prevent erosion (Fig. 4). Similar to traditional method, ACETube® helps to stabilize the coast. Moreover, the installation of ACETube is much easier and faster. The sea sand is used as the backfilling material inside the tubes, and create more space for large ships to berth and get access to Kaohsiung Port easily.

Location of ACETube®

Fig. 1. Location of jobsite

The water is 4 m in depth. In order to stabilize the cofferdam and coast lines, the pyramid-shape 2+1 ACETube® covered with ripraps is adopted. Furthermore, to decrease the tube q’ty, ACETube® (Type A) with circumference 36 m at the bottom and ACETube® (Type B) with circumference 32 m on top are adopted. Two layers of ACETube® can reach the designed height. 1.2

Design Concept

In addition to high cost and difficult construction, the traditional reinforced concrete method requires a lot of manpower and machines to complete the construction. With tradition method, although the purpose of protection and stability effects can be achieved, carbon emissions during the construction process are too high to achieve ecological friendliness. Considering the convenience of construction, energy conservation and carbon reduction, ACETube® requires sand pumping and small vessels for installation, and can be filled with existing sand source, which has the lowest impact to the environment. ACETube® demonstrates a flexible application method and helps silt treatment, resistance to scouring and ecological protection. There are two major functions that ACETube® plays in coastal erosion protection. One is the temporary cofferdam installed with ACETube® as the core of the embankment covered with ripraps to resist water erosion and protect the shoreline, and avoid the loss of the backfilling sand; the

Giant Geotextile Tube Applied to the Temporary Cofferdam Reclamation Construction

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other one is to fill the ACETube® with in-situ seabed sand to provide more space for large ships. Being cost-effective, eco-friendly, and functional, ACETube® is undoubtedly the best solution.

Caisson Outstretched

Fig. 2. The Caisson is outstretched

Backfilled with sea sand

Fig. 3. Backfilled with sea sand

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L-Shape cofferdam

Fig. 4. Location of ACETube®

The temporary cofferdams around the newly backfilled S4–S5 terminal of Kaohsiung Intercontinental Container Terminal are divided into east-west direction and north-south direction (Fig. 5). The designed height of the embankment is 7.5 m and the total length of the structure is 700 m. The purpose is to prevent the newly backfilled land from erosion during the rainy season or typhoon. Considering the local marine meteorological data, the structure stacked with the ACETube® is designed to be 2+1, the pyramid shape (Fig. 6). The bottom layer adopts ACETube® Type A (circumference 36 m), and the upper layer adopts ACETube® Type B (circumference 32 m). In addition, a layer of non-woven geotextile is laid on the surface as a separation layer. At last, the outer layer is covered with 300–500 kg riprap as surface protection. 1.3

Analysis Method

ACETube® for this project is fabricated from woven polypropylene geotextile with 250 kN/m of biaxial tensile strength. Per the analysis from software GeoCoPs, the setting of unit weight of sand is 16.5 kN/m, and underwater 4.7 m for filling of ACETube® TypeA. Under high modulus analysis and taking installation damage (e.g. pump pressure control), material durability in marine environment, creep and seaming strength etc. into consideration, the requirement of tensile strength of geotextile can be concluded as below Table 1. The suggested maximum filling height of ACETube® TypeA is 4.7 m. As to ACETube® TypeB, it is filled above water. According to analysis results, the maximum filling height is 3.8 m. Per below Table 1, it is easily to find out that ACETube® TypeB is analysed per installation above water. Without the

Giant Geotextile Tube Applied to the Temporary Cofferdam Reclamation Construction

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East-west Cofferdam

2+1 Pyramid-shaped stacking of ACETube® (Type A+Type B)

North-South Cofferdam

Fig. 5. L-Shaped Cofferdam

ACETube® Type B ACETube® Type A

ACETube® Type A

Fig. 6. Design cross drawing

confinement of water pressure, the requirement of tensile strength of geotextile is much higher. The results of ACETube Type A (Fig. 7), ACETube® Type B (Fig. 8), and size design parameters are as below (Table 1). Table 1. ACETube size design parameters (by GeoCops) ACETube® Type A ACETube® Type B Filling Height (m) – Hmax 4.7 3.8 Width (m) 15.7 14.2 Longitudinal ultimate tensile strength (kN/m) 111.8 169.7 Latitudinal ultimate tensile strength (kN/m) 137.4 204.2 Filling pump pressure(kPa) 1.2 1.5 Estimated filling height(m) 4.3 3.4

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Fig. 7. Results of underwater analysis (ACETube® Type A)

Fig. 8. Results of above water analysis (ACETube® Type B)

1.4

Installation Process

This project is a temporary cofferdam so it needs to anchor ACETube® both at sea side and coast size. For sea side anchoring – installation work is on a ship platform and using ropes to tie the loops of ACETube® to keep it in the proper place (Fig. 9). For coast side anchoring – Tie the ropes on the steel bars on the anchor on shore. The anchor consists of 3 bars of steel with total length 2 m. The depth buried underground is 1.5 m and the rest 0.5 m is exposed above the ground. Some of ACETube® are filled underwater (Fig. 10), so it’s necessary to remove all sharp objects which may cause damage to ACETube® before construction. After finishing cleaning the sea bed ACETube® must be anchored in the proper place and then filling begins. Successful filling depends upon suitable anchors. In the process of filling, stress of geosynthetic materials will change as filling pressure increases. That is why exceeding pressure may cause damage to geosynthetic materials. To control pump pressure is a key while filling. ACETube® has nonwoven geotextile lining which can keep more sand inside. However, to avoid exceeding pressure due to lower permeability with the nonwoven geotextile lining. The design of the chimney considers the pressure release and the height requirement, so the quantity, position and size need to be specially designed.

Giant Geotextile Tube Applied to the Temporary Cofferdam Reclamation Construction

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Fig. 9. Achoring installation of ACETube® (a) Sea side anchoring, (b) Coast side anchoring

Fig. 10. Filling construction of ACETube® (a) Cleaning sea bed, (b) Filling underwater

2 Conclusions ACETube® is one of flexible construction methods. It can be used for diversified applications. In comparison with the traditional method, ACETube® has the following advantages: – Filling materials are originated from jobsite so the construction causes less damage to local environment. – ACETube® is used for coastline protection instead of the traditional method. It provides an innovative and ecological application. – ACETube® with the advantages of high tensile strength, high permeability, and light weight brings the advantages of cost effectiveness, convenience, and short construction period in the process of installation. – ACETube® is used as a core of cofferdams covering with rocks. This way largely reduces the consumption of concrete and costs. – ACETube® helps not only protecting coastlines, but also for dredging in the harbor and generate more space to berth giant ships.

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References 1. GRI Test Method GT10. Test Method, Properties and Frequencies for High Strength Geotextile Tubes used as Coastal and Riverine Structures. Geosynthetic Institute, Kedron Avenue, Folsom, USA 2. GRI Test Method GT11. Installation of Geotextile Tubes used for Coastal and Riverine Structure. Geosynthetic Institute, Kedron Avenue, Folsom, USA 3. Jones, L.D., Davies, J.E., et al.: Geotextile tube structure guidelines for construct specifications. US Army Corps of Engineering (2006)

Effect of Strain Rate on Cyclic Behavior of Pond Ash Reinforced with Geocell Swaraj Chowdhury(&) and NiharRanjan Patra Department of Civil Engineering, Indian Institute of Technology, Kanpur 208016, India {swaraj,nrpatra}@iitk.ac.in

Abstract. This paper studies the effect shear strain on cyclic behavior of unreinforced and geocell reinforced pond ash samples. Pond ash was collected from upstream side of an ash embankment of Panki thermal power plant, Kanpur, India (seismic zone III). The size of pond ash samples was 70 mm
140 mm. According to the size of the sample, geocells were fabricated using high density polyethylene sheets with three interconnected cells of circular cross section. A series of consolidated undrained (CU) cyclic triaxial tests at a constant loading frequency of 1 Hz were carried out on pond ash samples with and without geocell reinforcement. These tests were carried out with varying cyclic shear strain rate (0.2%, 0.3% and 0.4%) over a constant confining pressure of 50 kPa. Results show that geocell reinforced pond ash exhibit better liquefaction potential and secant shear modulus than unreinforced pond ash samples. The secant shear modulus decreases with increase in cyclic shear strain rate from 0.2% to 0.4% for both unreinforced and geocell reinforced pond ash samples. The decrease is more pronounce in unreinforced pond ash samples. The number of cycles for initiation of initial liquefaction increases about 89% to 105% for geocell reinforced pond ash samples as compared to unreinforced samples. Keywords: Pond ash  Geocell  Sample size  Cyclic shear strain rate Liquefaction resistance  Secant shear modulus



1 Introduction Pond ash is a waste material which we get from thermal power plant. In India Thermal power plants are the main resource of power generation, these plants are using pulverized coal as fuel which generates huge amount of coal ash as a by-product. Dhadse et al. (2008) reported that annually about 112 million tons of ash is generated, from which 38 million tons are being used for land reclamation and construction work. Disposal of this ash is a huge problem as it requires huge amount of space and responsible for environmental pollution. In India, 26,325 ha of land has been used for coal ash disposal. To minimize this environmental and disposal problem coal ash and fly ash are mixed with water and deposited in ash ponds. This pond ash is a non-plastic material and possess very low dry density as compared to natural soil. To utilize pond ash as a geo-structural material in embankment construction particularly in earthquake prone areas it is important to improve its liquefaction behavior. © Springer Nature Switzerland AG 2020 F. Tatsouka et al. (Eds.): GeoMEast 2019, SUCI, pp. 9–21, 2020. https://doi.org/10.1007/978-3-030-34242-5_2

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According to Pandian (2004) fine grained ash particles are very prone to liquefaction. Dey and Gandhi (2008) performed stress controlled cyclic triaxial tests on pond ash concluded that pond ash in saturated condition becomes more liquefiable. Jakka et al. (2010b) performed experiments on pond ash samples collected from two different ash dyke and noticed that outflow ash samples possessed low shear strength. Jakka et al. (2010a) conducted stress controlled cyclic triaxial tests on these pond ash samples and found that the outflow ash samples were more liquefiable than inflow ash samples. Vijayasri et al. (2016) conducted a series of strain controlled cyclic triaxial tests on Renusagar pond ash samples reinforced with and without geotextiles considering the effect of loading frequency, number of reinforcing layers, strain amplitude and confining pressure. Geocells are three dimensional, polymeric, honey comb like structure of cells which are interconnected by joints. In recent past, these geocells are widely used for geotechnical engineering purposes like controlling erosion of slopes, enhancing bearing capacity of footing, reinforcing soft soils and slopes and protecting channel beds. Geocells are made from high density polyethylene (HDPE) with an open cell structure for containing soil or soil like material. Bathurst and Karpurapu (1993) conducted a series of large scale triaxial compression tests on 200 mm high soil samples reinforced with and without geocell and found that stiffness and strength was increased due to confinement effect geocells. Rajagopal et al. (1999) conducted a large number of triaxial compression tests on granular soil with single and multiple geocell confinement. The geocells were made using woven and nonwoven geotextiles and soft mesh to evaluate the effect of stiffness of geocell. It was found that granular soil develops a good amount of apparent cohesive strength due to geocell confinement and three interconnected cells is adequate to interpret the behavior of geocell reinforced layer which consists many interconnected cells. Shen (2005) carried out a series of triaxial compression tests to highlight the relative density effect of soil on strength parameters of two different aggregates reinforced with geocells. They concluded that peak friction angle and apparent cohesion of geocell reinforced soil increased with relative density. Wesseloo et al. (2009) performed a series of uniaxial compression tests on single cells and square shaped geocell packs of different sizes. They concluded that strength of geocell reinforced material was proportional to the size of individual cells. Considerable research has also been done in the area of geocell reinforced structures, such as shallow foundations (Dash et al. 2003; Sireesh et al. 2009; Pokharel et al. 2010; Moghaddas Tafreshi and Dawson 2010, 2012), slopes (Chen and Chiu 2008; Leshchinsky et al. 2009), pavements (Thakur et al. 2012; Leshchinsky and Ling 2013), and embankments (Zhang et al. 2010). From the past research work it may be concluded that study on the effect of cyclic shear strain rate on cyclic behavior of geocell reinforced pond ash samples is limited. In present study, experimental investigations have been carried out to understand the effect of strain rate on cyclic behavior of Panki pond ash samples with and without geocell reinforcement. The geocells have been fabricated using high density polyethylene sheets with three interconnected cells of circular cross section, the length of the geocell is equal to that of pond ash sample. The strain controlled cyclic triaxial (CU) tests have been

Effect of Strain Rate on Cyclic Behavior of Pond Ash Reinforced with Geocell

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performed with varying cyclic shear strain amplitudes (0.2%, 0.3% and 0.4%) at a confining pressure of 50 kPa. These tests have been carried out at constant loading frequency of 1 Hz. Based on these results the effect of strain rate on cyclic behavior of unreinforced and geocell reinforced pond ash samples has been discussed.

2 Materials and Methodology 2.1

Panki Pond Ash

Pond ash used for experiments collected from the upstream side of a pond ash dyke located inside Panki thermal power plant, Kanpur in India (Seismic zone III). Collection point of pond ash was 350 m away from the slurry disposal point that’s why the pond ash contains relatively finer size of particles. Initial investigations were carried out on Panki pond ash samples to find out its basic geotechnical properties. The physical properties of pond ash may vary with the depth and distance of the collection point from the slurry disposal point. 2.2

Geocell

Commercially available high density polyethylene (HDPE) sheet having thickness of 0.8 mm has been used to fabricate geocells. Tensile strength is 10 kN/m. Three cells made from HDPE sheet were stitched together into circular shape with a height equal to that of pond ash sample i.e. 140 mm to form the geocells required for the tests. Each cell are having diameter of 26 mm and height of 140 mm. The diameter of cells have been decided on the basis of trial tests which are carried out with varying diameter of cells. The thread and number of stiches used for attaching the cells was same throughout the testing program to have uniform condition for all the tests. Figure 1 shows the orientation of cells inside the geocell reinforced pond ash samples.

Fig. 1. Orientation of cells (26 mm diameter) inside Panki pond ash samples

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Testing Program, Procedure and Sample Preparation

The basic geotechnical properties of Panki pond ash like specific gravity, grain size distribution, compaction characteristics and consolidation characteristics have been carried out as per IS: 2720 Part III (1980b), IS: 2720 Part IV (1985), IS: 2720 Part VII (1980a), IS: 2720 Part XV (1965) standards. 70 mm
140 mm samples have been prepared at optimum moisture content to investigate the undrained cyclic behavior of compacted pond ash samples with and without geocell reinforcement. The cyclic triaxial (CU) tests with varying cyclic shear strain rate (0.2%, 0.3% and 0.4%) have been carried out to study the effect of strain rate on cyclic behavior of compacted pond ash samples with and without geocell reinforcement. All the cyclic triaxial (CU) tests have been carried out at 50 kPa confining pressure and 1 Hz loading frequency. These cyclic triaxial (CU) tests on pond ash samples with and without geocell reinforcement have been performed as per ASTM D5311-92 (ASTM 1992) and ASTM D3999-91 (ASTM 1996). Most of the tests were repeated twice to check the accuracy of test results. The Panki pond ash material mainly consists of silt size particles, presence of silt sized particles is more than 70% (i.e. 70.94%). Hence, moist-tamping technique suggested by Silver et al. (1976) is to be suitable for making compacted pond ash sample. The known mass of oven dried pond ash was mixed with an amount of water calculated from optimum moisture content. Then the mixture was placed in the 70 mm
140 mm split mold in four number of layers and every layer was relatively compacted by a tamping tool up to the required density. The density of each layer was checked after compaction, by measuring the height from top of the mold. After compaction each layer was scratched by knife for better interlocking. For geocell reinforced samples, first the geocell which was made by three interconnecting cells, placed at the center of the 70
140 mm split mold then the pond ash water mixture was put by using a spoon both inside and outside of the geocell in four layers and each layer was relatively compacted by a 25 mm diameter tamping rod to achieve the required density. The height of the geocell was equal to the pond ash sample height i.e. 140 mm and the center line of three interconnected cells was coinciding with the center line of the sample. So, the effect of confinement of geocell maintained throughout the length of the reinforced samples. Figure 1 shows the orientation of cells inside the pond ash samples. After that filter paper along with boiled porous stones were placed at top and bottom of the sample. Then the sample was covered by a rubber membrane and sealed using four O-rings. After that saturation tube was attached with upper base plate. Then saturation process was started using back pressure saturation system and it continued until B parameter reaches 0.9 which implies 90% saturation. Then the consolidation process was allowed at a particular effective confining pressure i.e. 50 kPa and the volume change was recorded during consolidation; after that shearing of sample was pursued without changing the effective confining pressure.

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3 Results and Discussions 3.1

Basic Geotechnical Properties of Panki Pond Ash

Panki pond ash mainly consists of silt-sized particles (70.94%), sand sized particles (26.22%) and clay sized particles (2.6%). The maximum dry density is 11.56 kN/m3 and its corresponding optimum water content is 30.86%. The value of coefficient of uniformity (Cu) was 8.83 and coefficient of curvature (Cc) was 0.49. The grain size distribution curve obtained from sieve and hydrometer analysis and compaction curve obtained from standard proctor compaction are shown in Fig. 2. The Coefficient of permeability value was 2.11
10−7 m/s and compression index was 0.046. The detailed basic geotechnical properties of Panki pond ash samples are shown in Table 1.

(a)

(b)

Fig. 2. (a) Compaction curve; (b) Grain size distribution of Panki pond ash Table 1. Basic geotechnical properties of Panki pond ash Physical Properties Specific gravity (G) Gravel size particles percentage (>4.75 mm) Fine sand size particles percentage (0.075–0.425 mm) Silt size particles percentage (0.002–0.075 mm) Clay size particles percentage (1200 >1600 246 840 647 35 180

5 Data Analysis The sheet of the geomembrane GEO-6 has the lowest SCR resistance (35 h) due to the fact that the polymer which it was manufactured is the one with the highest density 0,940 g/cm3, lowest MFR and because of that, a narrow weight distribution as seen in Fig. 4.

Fig. 4. Curves of molecular weight distribution of each polymer.

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If density would be the only or the most influential parameter in the SCR values, we wouldn’t justify that GEO-7, manufactured with a 0,941 g/cm3 density polymer had a SCR resistance value of 180 h higher than the one of GEO-6. The main reason for this result is that the molecular weight distribution is significantly wider for polymer PE-7. Polymers PE-1, PE-3 and PE-6 have a very similar weight distribution however the SCR resistance is very different. The reason is that the density of PE-6 is higher than the density of PE-3, and this higher than density of PE-1, respectively. So, for narrow molecular weights distribution the density limit has to be around 0,935 g/cm3, so the stress cracking resistance could be near 300 h. Polymer with higher densities could improve SCR if the molecular weight distribution is wider as GEO-4 (d = 0,937 g/cm3) and GEO-5 (d = 0,938 g/cm3). Maybe the best situation if for PE-2 with a density of 0,934 g/cm3 and the second wider molecular weight distribution, presents a SCR more than 1000 h.

6 Conclusions There is no direct relationship between the polymer density and stress crack resistance. Also, there is no evidence of relationship between the molecular weight distribution and its stress crack resistance. Polyethylenes with higher density but a wider molecular weight distribution, improve the stress crack resistance values. Polyethylenes with a narrow molecular weight distribution but lower density, improve their performance in terms of stress crack resistance. One of the worst situations occurs for polyethylenes with densities  0.940 g/cm3 and narrow molecular weight distribution. The best situation will be found in lower-density polyethylene and wider molecular weight distribution until the own process allows it or the mechanical or chemical geomembrane’s properties.

References 1. Hsuan, Y.G.: Data base of fields incidents used to establish HDPE geomembrane stress crack resistance specifications (1999) 2. Scheirs, J.: A guide to polymeric Geomembranes (2009) 3. CSIC. Ciencia y Tecnología de materiales poliméricos (2004) 4. Koerner, R.M., Hsuan, G.Y., Koerner G.R.: Geomembrane lifetime prediction: unexposed and exposed conditions. Geosynthetic Institute (2005)

A Geomembrane Liner to Stop Water Seepage in an 8 km Long Embankment: The Bill Young Reservoir Case History Alberto Scuero1(&), Gabriella Vaschetti2, and John Wilkes3 1 Carpi Group, Balerna, Switzerland [email protected] 2 Carpi Tech, Balerna, Switzerland [email protected] 3 Carpi USA, USA [email protected]

Abstract. The C.W. “Bill” Young Regional Reservoir in USA, owned and operated by Tampa Bay Water, holds 58,673,855 m3 of raw water and is a vital resource for 2.4 million residents of Tampa Bay area. During the wet season, water from surrounding rivers is stored for withdrawals during the dry season greatly reducing the dependence on Florida’s waning groundwater. The reservoir is formed by an approximately 8 km long earthen ring dam of 15 m average height. The reservoir started operation in 2005. Unusual cracking of the reservoir’s upstream soil-cement erosion protection system was observed about one year after the first fill. The reservoir was taken offline in 2012 to prepare for the renovation of the upstream slope. The designer, Gannett Fleming, after subsurface investigations and an extensive design process, selected an upstream solution with a 50-year design life. The rehabilitation included removing the original high-density polyethylene liner, replacing it with a polyvinylchloride liner, and constructing a robust soil-cement erosion protection system placed on a gravel drainage. The selected liner, SIBELON® 2CNT 3300, is a 2 mm thick geomembrane with a geotextile layer heat-bonded on both sides of the geomembrane. The geotextile layer provides the required friction angle with the earthen embankment below and the gravel drainage layer above. Overlying the liner, the soil-cement was designed to provide a minimum of 50 years of erosion protection. The renovation started in February 2013 and the reservoir was placed back into operation by the end of 2014. Over the following years the reservoir has been frequently monitored for dam safety. This paper explores aspects of the design and construction of the renovation features, and presents the dam safety monitoring findings to date, including piezometric water levels in the soilcement erosion protection system and the seepage flow observed surrounding the reservoir.

1 Introduction Originally constructed from 2003 to 2005, the 15.5 billion gallon C. W. “Bill” Young Regional Reservoir located just east of Tampa, Florida, was built to supply water to Tampa Bay Water’s customers. Water is captured from the surrounding rivers and © Springer Nature Switzerland AG 2020 F. Tatsouka et al. (Eds.): GeoMEast 2019, SUCI, pp. 68–81, 2020. https://doi.org/10.1007/978-3-030-34242-5_7

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stored within the reservoir in an effort to reduce the area’s dependence on Florida’s waning groundwater. The reservoir is surrounded by approximately 8 km of earthen embankment. The original design included an about 40 cm thick flat-plate soil-cement erosion protection, underlain by a 1.5 mm high-density polyethylene (HDPE) geomembrane, which was tied into a soil-bentonite foundation cutoff wall at the upstream toe. First filling of the reservoir was conducted in November 2005. By the end of 2006, the second drawdown cycle revealed unusual cracking in the soil-cement. Tampa Bay Water performed numerous investigations and repairs to the soil-cement with no success. In 2012, Tampa Bay Water took the reservoir offline and began the $129 million rehabilitation construction that is object of the paper.

2 Reservoir Rehabilitation The design-build team of Kiewit Infrastructure South, Inc. and Gannett Fleming, Inc. was selected to design and construct a reservoir that could provide the storage of raw water for the region, with minimal maintenance throughout its 50-year design life. As part of the rehabilitation, the entire upstream slope of the embankment dam was demolished and re-constructed with a robust and resilient soil-cement erosion protection and seepage barrier system. The previous soil-cement and HDPE geomembrane were removed and substituted by an about 60 cm thick stair-stepped soil-cement erosion protection, a gravel drainage layer about 20 cm thick, and a seepage barrier consisting of a geomembrane formulated with a special compound of polyvinylchloride plasticized with high molecular weight branched plasticizers (Fig. 1).

Fig. 1. Typical cross-section of the reservoir’s rehabilitated upstream slope

2.1

Waterproofing Geocomposite Liner

The waterproofing liner is a composite membrane (geocomposite). The watertight element is a SIBELON® geomembrane formulated with a special compound of polyvinylchloride plasticised with high molecular weight branched plasticisers, which

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has been used since 1980 on large dams, reservoirs canals, hydraulic tunnels and shafts worldwide, and is well documented in international literature (ICOLD 2010, GeoMeast 2018). Due to the complexity and specific requirements of the project, Carpi created a new product, SIBELON® 2CNT 3300, that could meet the required friction angles for both the earthen embankment below and the gravel drainage layer above, as well as provide cushioning from the placement of the gravel drainage layer. To meet these requirements, two geotextiles were heat bonded to the SIBELON® geomembrane during extrusion. The bottom geotextile was a 200 g/m2 polyester geotextile, while the top layer was a 500 g/m2 polypropylene geotextile (Fig. 2). This new geocomposite was used to cover the more than 370,000 m2 of slope area along the full 8 km length of embankment dam. This translated into nearly 4,000 rolls of approximately 2 m wide geocomposite.

Fig. 2. SIBELON® 2CNT 3300 geocomposite liner

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Upstream Design Considerations and Features

2.2.1 General Considerations As the cracking observed in the original soil-cement erosion layer was the impetus for the rehabilitation project, seepage and stability analyses were conducted to assess the new erosion protection system design. Seepage and slope stability cross- sections were developed based on the extensive on-site subsurface investigation and different embankment configurations. The subsurface investigation included numerous field and laboratory testing to characterize the soils and to determine the design parameters. Since the rehabilitation of the upstream slope did not affect the downstream slope and the overall seepage surrounding the reservoir, the seepage analysis performed for the rehabilitation project was used to determine the phreatic surface behind the new geocomposite and within the gravel drainage layer. This analysis was used to calculate seepage forces behind the geocomposite and to check the uplift resistance of the stairstepped soil-cement.

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2.2.2 Veneer Slope Stability Stability of the new waterproofing liner on the prepared embankment upstream surface was evaluated as part of the design. The friction angle between the geocomposite membrane and the underlying embankment was identified as the most critical in the suite of interfaces that exist in the geocomposite and surrounding design features. A soil-cement buttress was constructed at the toe of the upstream slope (Fig. 3) to provide support to the stability of the overall erosion protection system; however, this buttress was ignored in the veneer slope stability analysis for conservatism.

Fig. 3. Layering and toe detail

To understand the friction angles developed in the various materials in the crosssection shown in Fig. 3, laboratory testing was conducted. The veneer slope stability analysis was performed by Dr. Robert Koerner and documented in the report entitled “Slope Analysis of Tampa Bay Reservoir Renovation Project.” Using the interface friction angles taken from laboratory testing, the minimum interface friction angle that could exist in the veneer of materials was 27.9°. The post peak friction angle for the geotextile heat bonded to the PVC membrane falls below this value at 25.5°, but the method by which the geotextile is bonded to the PVC membrane essentially precludes the ability of the geotextile to de-bond from the PVC membrane. Dr. J. P. Giroud reviewed the veneer slope stability analysis of the geocomposite system for Carpi. An infinite slope failure was modeled to conservatively estimate the interface friction between the geocomposite membrane and upstream embankment face. This value was used to establish the criteria for required minimum friction angle and determined to be 29.5°. All of the friction angles were found to be at least 29.5° and found to be stable at the 2.5H:1V upstream slope. 2.2.3 Interface Stability of Geosynthetic Layers To our knowledge, this was the first time in the world that a double geotextile, coupled to a PVC geomembrane, had been produced in large quantities for a hydraulic structure. After the friction angles were developed for the geocomposite system on the slope and gravel layer, the critical parameter for the geocomposite production was the adhesion between the geotextiles and the PVC geomembrane. A healthy tolerance was placed on the value to ensure the manufactured material had the geotextile stoutly

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bonded to the PVC geomembrane. This was monitored carefully throughout the production. However, due to the strong adhesion of the top polypropylene geotextile to the PVC geomembrane, the two could not be easily separated in the field for overlapping adjacent rolls around the curves. This challenge eventually led to the team to develop a different approach to avoid stripping the geotextile for welding rolls in the field. During the design, it was apparent that the installation of the PVC geocomposite would be difficult in the curves. In order for the top geotextile to withstand the placement of the soil-cement over the gravel drainage layer, the geotextile had to be resistant to deterioration from cement. Polyester geotextiles deteriorate when they come in contact with cement and thus were not acceptable as the top geotextile of the geocomposite. As such, a polypropylene geotextile was required. However, to install a watertight seepage barrier along the embankment alignment curves, the polypropylene geotextile had to be removed to allow for bonding between two adjacent rolls of geocomposite. As such, adjoining rolls were designed with offsetting exposed weld strip areas (Fig. 4).

Fig. 4. The roll on the left was the primary geocomposite, Sibelon CNT® 3300, a PVC geomembrane with polyester geotextile on the bottom towards the slope and polypropylene geotextile on the top face, with weld strips on opposing longitudinal corners. The right diagram is the geocomposite Sibelon CNT® 2800 with PVC geomembrane and polyester geotextile on the bottom face of the slope with an exposed face to allow for overlapping and welding in curves

The veneer slope stability analyses determined the necessary minimum area of geotextile needed to provide overall stability to the embankment. The analysis determined that every third roll could be free of geotextile on the top face, while maintaining proper factors of safety for stability. This resulted in adding two actions to the installation: (1) for each curved area, the crew must be given very precise overlap dimensions at the crest for the rolls to lineup at the toe and (2) after the rolls are welded together, a polypropylene geotextile had to be spot welded to the exposed PVC geomembrane face to protect against the gravel drainage layer above (Figs. 4 and 5). Further analysis was performed to ensure that this alternating roll layout would still meet the required friction angle for the drainage layer above. While some portions of the geotextile were not originally heat bonded to the geomembrane, there was a high enough proportion of geotextile along the length of the curve to meet the required friction angle.

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Fig. 5. The photograph on the left shows how two rolls of full double-faced geotextile geocomposites were placed in a curve and then a roll of geocomposite without the geotextile was placed too allow overlapping in a curve. The photograph on the right shows how a geotextile is then spot welded over the exposed PVC to protect it from the rock drainage layer

2.2.4 Global Stability and Longevity Global stability of the upstream slope was evaluated for the end of construction condition, long-term steady-state seepage condition, and the rapid drawdown condition (both operational drawdown and emergency controlled release). In the original design, the upstream slope was construction on a 3H:1V slope, with a geomembrane below the soil-cement, placed on a 2.5H:1V slope. The slope stability analyses determined the final slope of 2.5H:1 V was stable and the design could eliminate the need to return the interior of the reservoir to a 3H:1V slope. Additionally, a soil-cement toe buttress was also analyzed and included in the final design to prevent sliding of the soil-cement stair-steps. The longevity of the soil-cement was also a critical part of the design. The stairstepped design allowed for easier construction and would prevent the delamination of lifts, thus improving long-term durability. The minimum required 7-day compressive strength was set at 4.48 MPa and a maximum 14% weight loss was required for the wet-dry durability testing. The longevity of the soil-cement would not only be dependent upon the design, but also upon the construction methods. As the soilcement was to be placed parallel to the slope, the area largely susceptible to weathering would be the stair-steps themselves, most likely due to direct wave impacts and frequent wetting and drying cycles. Achieving compaction along the lift, adjacent to risers of the steps would be difficult and require care quality control during construction. Prior to determining the soil-cement mix, the on-site soils were evaluated to determine if they were suitable for use. The soils predominately found on-site were classified as silty sand (SM) or poorly-graded sand with silt (SP-SM) and were tested for low pH, excessive organics, and excessive clay content. Based on these tests, the soil was found to be acceptable for use in the soil-cement mix design. A test batch program was conducted to determine the appropriate cement content for the mix to achieve the required compressive strength and durability requirements. A soil-cement batch plant was erected within the interior of the dewatered reservoir. Stockpiles of the on- site soil were placed adjacent to the batch plant in horizontal lifts and were excavated from the side of the piles to help mix the soil aggregate. The soil was then placed through screens to sift out particles greater than ½-in. in diameter and concentrations of clay.

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2.2.5 Crest and Wave Barrier The renovation design also evaluated the wind and wave conditions, particularly tropical storms and hurricanes likely to pass over the region. While stair-stepped soilcement is typically intended to reduce or eliminate wave run-up during storm events, the stair-stepped design at this reservoir was not. A wave barrier wall was placed on the crest of the embankment and was designed to prevent wave-induced overtopping by returning any run-up back to the interior of the reservoir. To be conservative, the design of the wave barrier assumed a flat-plated soil-cement upstream slope (however, as previously mentioned, for ease of construction, the soil-cement was placed in stair-step lifts). As such, the presence of the stair-steps further reduces wave run-up, and provides additional erosion protection, in excess of the flat-plate’s 50-year design life. The wave barriers were affixed to the crest with mechanical anchors. The soil-cement was determined to be insufficient to properly anchor the barriers due to the low-strength of the material. As a result, the top three soil-cement stair-steps were replaced with roller compacted concrete. The wave barrier innovation is seen in only a handful of reservoirs across the United States. 2.3

Drainage Layer

While the thicker stair-stepped soil-cement provides additional weight against potential uplift forces, a drainage layer was designed and placed as the main line of defense against water pressure build-up below the soil-cement. The drainage layer consists of AASHTO No. 57 aggregate, benched by aggregate-filled gabion baskets (Fig. 6). The drainage layer was designed to have approximately 1,000 times more drainage capacity than the surrounding soil, thus providing a free draining system during filling and drawdown cycles of the reservoir.

Fig. 6. Aggregate-filled gabion baskets below the soil-cement erosion protection system

Uplift was calculated across the embankment and was determined to be detrimental if there was an approximately 1.2 m differential head across the soil-cement. In addition to the design considerations for the drain during reservoir operations, the design also considered the capacity necessary to pass a design storm during construction, when the drainage layer was likely to be exposed to the environment.

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Placement of the Cover Layer

After batching of the soil-cement, it was hauled by truck to the upstream slope and placed with one of the largest tracked conveyor systems ever used on a US dam (Fig. 7). Spreading of the mix was performed by dozers, in compacted lifts between 1.8 and 3.6 m thick. A bonding slurry was used between lifts to ensure proper adhesion of the subsequent lifts.

Fig. 7. Placement of the soil-cement via tracked conveyor syste

Compaction was achieved through the use of smooth drum vibratory rollers (Fig. 8). As previously noted, proper compaction of the soil-cement was a main concern during the design and extensive quality control was required during construction. Nuclear density tests were performed on the soil-cement in-situ and pills were molded and tested for compressive strength. Additionally, great care was taken along the stairstep edges; to ensure the steps were fully compacted, the soil-cement was overbuilt, compacted, and then the overbuild was then removed (Fig. 8).

Fig. 8. Compaction of the soil-cement lifts (left) and compaction along the stair-step edges with removal of the overbuild material (right)

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In all, > 401,390 m3 of soil-cement were batched, placed, and compacted along the > 7,900 m of embankment (Fig. 9).

Fig. 9. Stair-stepped soil-cement erosion protection system

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Installation Methods

Since the geocomposite material was to be covered by the drainage gravel and then the soil-cement, the only required anchorage points were along the crest and the toe. This anchorage was achieved by the use of trenches which were shaped in such a way to provide enough surface friction to hold the material in place, while minimizing trench excavation. Along the crest, the trench was dug following the earthwork preparation of the slope surface. A geocomposite roll was then partially unraveled and placed into the trench. Particular care was taken to ensure the polyester geotextile was in intimate contact with the earthen embankment below. The unraveled length of the geocomposite was held in place with a wood stake (Fig. 10). There are two types of curves along the embankment: concave and convex. At concave curves, in order for a constant width roll to be deployed, subsequent rolls must start with an overlap of just the width of the weld strips and be overlapped as the roll is deployed down the slope to take up the excess material in the curve. For a convex curve, the opposite is true as the rolls must be overlapped at the crest and then overlap gradually tapered down the slope towards the toe until the overlap is just in the weld strip area at the toe. Thus, making the roll geometry is very important. In addition to the alignment curves, the layout of the geocomposite rolls had to be adjusted to accommodate interior ramps. At the ramp locations, some of the geocomposite rolls had to be cut and reoriented with the varying geometry of the ramp. This adjustment primarily occurred at the end of the ramps where the grade changed neared the toe. Along some of the flat portions of the ramp the geocomposite was able to accommodate the change in grade and be deployed in one piece all the way to the toe.

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Fig. 10. The crest anchorage trench dug using a specially designed bucket to match the designed shape of the trench (left). Once a portion of trench digging was complete, geocomposite rolls could be inserted into the trench and deployed down the slope (right)

To create a continuous seepage barrier, the geocomposite rolls had to be heat welded together. The majority of these welds were completed using a dual track heat weld between overlapping rolls along a weldstrip (Fig. 11). These weldstrips were created by leaving the geotextiles off of the very edge of the roll width where overlapping of adjoining rolls would occur. In some locations with specific details, such as along the bridge columns, these welds were done with a manual heat welder.

Fig. 11. Dual track heat weld using a Twiny T Leister automatic welder (left). Multiple heat welds were started at once, to achieve high productivity (right)

Welding was completed starting in the crest anchor trench and progressing down the slope to the toe trench. Every weld required quality control checks performed using nondestructive and destructive methods to ensure a successfully completed weld was achieved. If a location failed any of the quality control checks, the weld would be analyzed in order to find the faulty location. Once found, the weld would be repaired and retested until a passing result was achieved. The design of the seepage barrier required a watertight transition from the geocomposite to the in-situ seepage cutoff wall. To create a watertight seal, only a geomembrane could be used in the trench. A SIBELON® C 2600 geomembrane was welded to the bottom of the geocomposite rolls that had been deployed along the

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embankment slope. Once all of the geosynthetic material was placed, the trench was backfilled with soil-bentonite (Fig. 12 at left).

Fig. 12. The geomembrane was installed in the toe trench then backfilled with a soil-bentonite mixture (left). The submersible watertight mechanical seal at a structure along the reservoir (right)

Watertight seals were also required where concrete structures tied into the soilcement (Fig. 12 at right). The seal consisted of a stainless-steel batten anchored in placed, above the geocomposite using threaded bars and chemical anchors. An EPDM gasket was placed between the batten and the geocomposite. Additionally, between the geocomposite and the concrete, a two-part epoxy resin was used to help to create a smooth surface, sealing the connection through pressure. There were two primary site conditions that combined to make the optimization of the installation an iterative process. First, typical summer weather in Florida includes daily afternoon thunderstorms. Secondly, the reservoir embankments were constructed of highly erodible soil (predominately sand). It was difficult to prepare large areas of the slope in an effort to maximize earthwork production. Additionally, the geocomposite absorbs a small amount of water by diffusion; if the geocomposite is allowed to sit and be soaked, it becomes more difficult and time consuming to weld. As a result, the installation became an iterative process to determine the correct ratio of time required for each operation—earthwork and geocomposite placement. It was important that the production be scheduled to allow all work to be completed before each daily rainstorm or forecasted weather event. Production had to be adjusted each day but increased once the proper balance was achieved. The construction of the rehabilitated upstream slope was completed in mid-2014, with full operation of the reservoir commencing by the end of 2014 (see Fig. 13).

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Fig. 13. C. W. “Bill” Young Regional Reservoir near completion of the upstream rehabilitation construction project (Photo taken September 2, 2014)

To monitor the seepage forces behind the geocomposite, piezometers were installed under the geocomposite at varying elevations (Fig. 14) around the reservoir. These piezometers below the geocomposite are continuously monitored and, to date, have not exhibited any seepage forces that appear to be detrimental to the upstream erosion protection system.

Fig. 14. Tampa Bay Water is currently performing a drawdown of the reservoir, allowing for increased visual inspection of the soil-cement slope protection

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Fig. 15. Configuration of the piezometers above and below the geocomposite

To assess the permeability of the drainage layer below the soil-cement, several piezometers were also placed above the geocomposite around the reservoir. These piezometers were tipped approximately three feet above the aggregate-filled gabion baskets at the toe of the soil-cement (see Fig. 14). As expected, all of these piezometers above the geocomposite have recorded water elevations nearly equivalent to the reservoir’s pool elevation. An example of one the piezometers monitoring the drainage layer permeability is shown in Fig. 15. As seen in this graph, the water elevation recorded by the piezometer follows the water elevation recorded within the reservoir.

3 Conclusions Gannett Fleming performed as the Permit Compliance Engineer for five years post construction with responsibilities including performing monthly and annual dam safety inspections of the reservoir. As part of the dam safety inspections, the upstream soilcement slope protection was visually inspected, and data were collected for monitoring the permeability of the drainage layer and the seepage through the embankment and foundation soils. To date, the soil-cement erosion protection system, gravel drainage layer, and geomembrane are performing as designed. As seen by the reservoir stage shown in Fig. 16, Tampa Bay Water perform a drawdown of the reservoir from elevation 136.5 feet NGVD29 to approximately elevation 110.0 feet NGVD29, a withdrawal of approximately 10 billion gallons. The drawdown has allowed the dam safety inspections to observe a larger portion of the soil-cement. Only minimal erosion and typical shrinkage cracking has been observed in the soil-cement, with crack widths of no more than a few millimeters. In the 5 years that followed installation, Gannett Fleming continued to monitor the embankment dam and rehabilitated upstream erosion protection system. The successful drawdown and continual monitoring have proved the reliability of the design as a longterm solution. By ensuring adequate water supply and drought resistance, the rehabilitated reservoir minimizes reliance on groundwater for customers today and for the Tampa Bay area’s future generations.

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Fig. 16. An example piezometer data monitoring the drainage layer below the soil-cement erosion protection

References ICOLD, International commission on Large Dams: Bulletin 135, Geomembrane Sealing Systems for Dams – Design principles and review of experience. ICOLD, Paris (2010) Scuero, A., Vaschetti, G.: Rehabilitation of Dams with Watertight Geomembranes, in the Dry and Underwater. GeoMEast 2017 International Congress (2018)

Experimental Research and Application of Geopolymer in Soft Soil Foundation Treatment Jialiang Yao, Haojie Qiu(&), Hua He, Xin Chen, and Guiyu Hao School of Traffic and Transportation Engineering, Changsha University of Science and Technology, Changsha, Hunan 410114, China [email protected]

Abstract. In order to understand the characteristics of geopolymer used as the curing agent to reinforce soft soil foundation by the mixing pile method, the unconfined compressive strength test and direct shear test were carried out in laboratory. The orthogonal test with three indexes of geopolymer content, soft soil moisture content and stirring time was carried out. The optimum mix proportion of geopolymer used for soft soil foundation treatment was analyzed by the range analysis. The mechanical properties of geopolymer stabilized soil were studied experimentally. From the test results, it is shown that the primary and secondary factors affecting reinforcing soft soil are geopolymer content and moisture content of soft soil. The optimum geopolymer content is 14%. The unconfined compressive strength and shear strength of the geopolymer stabilized soil at 90 days can reach 3.75–4.73 MPa and 0.753–1.523 MPa, respectively. The relationship between the compressive strength (UCS) and shear strength is s ¼ 0:26fcu . The performance of the soft soil stabilized with the geopolymer is better than that of the cement stabilized soil. Keywords: Geopolymer Application

 Soft soil  Orthogonal test  Microstructure 

1 Introduction The infrastructure construction on soft soils is highly risky due to its poor shear strength and high compressibility [1, 2]. Therefore, it is important to enhance the soil properties using stabilization techniques that can respond to increasingly demanding situations. In situ deep mixing is an effective means, which has been developed over two decades primarily to effect columnar inclusions into the soft ground to transform soft ground to composite ground. The deepmixing technology was simultaneously developed in Sweden and Japan using quicklime as a hardening agent. Subsequently, ordinary Portland cement slurry was used as a cementing agent because it is readily available at reasonable cost [3–5]. However, for some soils, the effect of cement reinforcement is not obvious, which limits the application scope of this method. In addition, this method needs a lot of cement. The production of cement consumed a large amount of coal and released a large amount of gas, such as CO2 and SO2 , which was unfavorable to resources and the environment [6–8]. Apart from the environmental © Springer Nature Switzerland AG 2020 F. Tatsouka et al. (Eds.): GeoMEast 2019, SUCI, pp. 82–94, 2020. https://doi.org/10.1007/978-3-030-34242-5_8

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drawbacks, ordinary Portland cement often shows a high plastic shrinkage and a reduction of mechanical strength due to the loss of water and incomplete hydration at early ages [1, 9, 10]. The term “geopolymer” was firstly introduced by Davidovits, who developed a binder obtained from the polycondensation of metakaolin in an alkali medium at the temperature of 100–150 °C [11]. As the raw material is mainly of geological origin (clay, kaolin), Davidovits termed geopolymer as a mineral polymer obtained by chemical processes of geosynthesis [12]. According to Provis and Deventer, geopolymers are framework structures produced by the condensation of tetrahedral aluminosilicate units, with alkali metal ions balancing the charge associated with tetrahedral Al [13, 14]. Geopolymers show advantages in immobilizing low- and intermediate-level nuclear waste [15–17]. Geopolymer concrete reduces the environmental damage that carbon dioxide, which is abundantly produced in the manufacturing of Portland cement, causes to the environment. Moreover, they also possess a compressive strength similar to that of cement-based materials, which explains the primary application of geopolymers as an alternative material to ordinary Portland cement [18], thus arousing the interest of researchers. Several properties of geopolymers have been investigated in the literature: chemical durability (acid attack, sulfate attack, alkali silica reaction, carbonation….) [18–20], fire-resistance [21] and electrical properties [18, 22, 23]. Guo Xiaolu et al. studied the dissolution polymerization mechanism of fly ash under the action of alkali activator and the macro-compressive strength, micro-structure and morphological characteristics of fly ash-matrix geopolymer after the formation. Then, the roles of alkali, water, silica-aluminium and calcium components in the reaction mechanism of fly ash-matrix geopolymer is expounded [24, 25]. Ding et al. used blast furnace slag and fly ash as raw materials. The inorganic polymer was used as a repairing material for concrete structures. The basic high viscosity and good working ability of blast furnace slag and fly ash were utilized. Based on the modulus(SiO2/Na2O)of activator,the optimum replacement ratio of geopolymer was determined. From the compressive strength test of concrete base course bonded with the geopolymer slurry, it is shown that repair rate of 120% can be achieved. By comparison with Portland cement, it can be proved that the slag-fly ash-matrix polymer slurry has a good potential for further engineering development in the future [26]. Sudipta Naskar et al. studied the effect of incorporation of nanomaterials, such as silica, carbon nanotubes and titanium dioxide on the compressive strength of geopolymer concrete in geopolymer concrete. From the test results, it is shown that a small number of nanomaterials (about 1%) can significantly improve the compressive strength [27]. Neha P Asrani et al. hybridized 5D hook-shaped end steel, polypropylene and glass at a certain amount. It was found that the incorporation into geopolymer composites led to an increase in impact property and ductility [28]. Kuo et al. used Fly ash/groundgranulated blast-furnace slag geopolymer (FGG) as the repair material for concrete with ground-granulated blast-furnace slag (GGBFS) as the main cement material. The slant shear strength and split tensile strength results indicated that the bond strength of the FGG repair material decreased with an increase in the liquid–solid ratio, and a favorable repair effectiveness could be achieved when the fly-ash substitution rate was 10% [29]. The geopolymer has a wide range of raw materials. It has low energy consumption, low environmental pollution, easy preparation, good mechanical

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properties and corrosion resistance. It is an environmentally friendly new building material with a wide application prospect [30–32]. The purposes of this paper are to study the physical and mechanical properties of soft soil foundation treated by geopolymers through indoor and outdoor experiments, and to develop a new soft foundation treatment method that can reduce the environmental pollution and be used in practical engineering with obvious effects. Therefore, the Wenzhou soft soil was taken as the research object. The reinforcement test of soft soil was carried out in the laboratory by using geopolymer. After 7 d, 28 d and 90 d reinforcement and maintenance of geopolymer stabilized soil, its mechanical properties indexes were determined by the direct fast shear test and the unconfined compressive strength test. The research results of this paper can provide a new method for soft foundation treatment.

2 Raw Materials and Orthogonal Test 2.1

Experiment Materials

In this paper, the DW-type geopolymer was used as the curing agent. The soft soil of the 104 National Road Reconstruction and Extension Project in Wenzhou, Zhejiang Province was used as the soil sample. 42:5# cement and water were used as auxiliary materials for research. The technical properties of the main raw materials are as follows: 2.1.1 DW-Type Geopolymer The DW-type geopolymer is an inorganic macromolecule material with gel structure. It is formed by using industrial waste residue, mineral active ingredients and other materials catalyzed by the alkali activator. It is a high performance inorganic cementitious material with excellent properties of organic polymer, ceramics and cement. Its physical and mechanical properties were shown in Table 1. Table 1. Indexes of physical and mechanical properties of DW-type geopolymer No.

Inspection items

1 2

Bleeding rate setting time

Unit of measurement % min

Cement standard 5 3% Initial setting time = 45 Final setting time 5 600 = 55

Measured values of DWtype geopolymer 0.5% 165 243

28d flexural MPa 8.7 strength 4 28d MPa = 32:5 52.1 compressive strength Note: At present, the properties indexes of geopolymers have no standard. Therefore, it refers to the technical standards of Portland cement. 3

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2.1.2 Soft Soil Sample The brown muddy clay was selected as the soft soil sample. It was crushed and dried in oven after transported from the site. After dried, it was screened by 5 mm screen and stored in sealed barrel. The physical properties of soft soils were shown in Table 2. Table 2. Indexes of physical properties of soft soil sample Testing items

Sampling depth m

Natural moisture content xð%Þ

Test results

9.0

52.0

2.1.3

Wet density q ðg=cm3 Þ 1.60

Pore ratio e

Plasticity index Ip

1.463 28.5

Compression Organic factor ðMPaÞ matter ð%Þ 1.41

2.11

Other Materials

(1) Cement: 42:5# ordinary Portland cement was provided by a cement plant in Changsha. The physical and mechanical properties were shown in Table 3. (2) Water: drinking water Table 3. Indexes of physical properties of cement Standard consistency water consumption (%)

0.08 mm sieve residue rate (%)

26.1

2.1

2.2

Setting time (min)

Stability

Initial setting 250

Final setting 390

–

Flexural strength (MPa) 3d 28d

3d

28d

qualified

4.1

24.2

48.3

6.8

UCS(MPa)

Test Scheme

The orthogonal design method was proposed to design a three-factor and three-level orthogonal test. The influences of geopolymer content, moisture content of soft soil and stirring time on the strength of geopolymer stabilized soil were analyzed. The optimum geopolymer content and the optimal horizontal combination were obtained. The unconfined compressive strength (UCS) and the shear strength of specimens at 7 d, 28 d and 90 d were measured. The UCS is the average value of three specimens. The factors and levels were shown in Table 4. The orthogonal table L9 ð33 Þ was used in this experiment. The specific orthogonal test scheme was shown in Table 5.

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Soft soil moisture content xð%Þ 46 49 52

Stirring time (s) 240 480 720

Table 5. Orthogonal test scheme Level Test factors Geopolymer content aw ð%Þ 1 10 2 10 3 10 4 12 5 12 6 12 7 14 8 14 9 14

Soft soil moisture content xð%Þ 46 49 52 46 49 52 46 49 52

Stirring time (s) 240 480 720 240 480 720 240 480 720

2.2.1 Specimen Preparation G and Testing The geopolymer contents in this test are 10%, 12%, 14% respectively. Considering the influence of water content of soft soil on the strength of the geopolymer stabilized soil, the water content of geopolymer stabilized soil must fluctuate about 3% according to its natural water content. And the water-cement ratio is 0.5. According to the determined mix proportion, the size of cube test model (70.7 mm  70.7 mm  70.7 mm) and ring sampler (diameter 61.8 mm and height 20 mm), the suitable geopolymer powder, soil sample and water were taken to prepare the test specimens of compressive strength and shear strength respectively. After 1–2 d, the specimens with a certain strength were demoulded. The specimens were placed in a curing box for curing. The curing temperature of specimens is 20 ± 2 °C, and its humidity is maintained at about 75%. The unconfined compressive strength test and shear strength test were carried out on the specimens after 7–90 d curing in the curing box. 2.3

Orthogonal Test Results and Analysis

2.3.1 Orthogonal Test Results The UCS and shear strength of geopolymer stabilized soil specimens at 7 d, 28 d, and 90 d were shown in Table 6. The range analysis of test results in Table 6 was carried out. The calculated results were shown in Tables 7 and 8. The trend chart was shown in Fig. 1. In Tables 6, 7 and 8, A is the geopolymer content; B is the soft soil moisture content; C is the stirring time.

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Table 6. L9(33 ) Orthogonal test design and results No. A B 1 2 3 4 5 6 7 8 9

10 10 10 12 12 12 14 14 14

46 49 52 46 49 52 46 49 52

C 240 480 720 480 720 240 720 240 480

UCS/ MPa 7d 28d 0.56 1.95 0.53 1.86 0.51 1.70 0.72 2.24 0.64 2.15 0.58 1.99 0.89 2.48 0.84 2.35 0.77 2.27

90d 3.99 3.87 3.75 4.20 4.11 4.02 4.73 4.39 4.25

Shear 7d 0.792 0.187 0.170 0.237 0.223 0.211 0.265 0.257 0.242

strength/MPa 28d 90d 0.512 0.778 0.442 0.765 0.345 0.753 0.629 1.239 0.597 1.145 0.532 1.043 0.742 1.523 0.726 1.447 0.641 1.314

2.3.2 Range Analysis of Orthogonal Test From Tables 7 and 8, it can be concluded that the compressive strength and shear strength of the geopolymer stabilized soil specimens basically have similar change law, and the order of factors affecting the strength of specimens is A > B > C. Table 7. Range analysis of various factors on UCS Range

7d UCS A B K1j 1.60 2.17 K2j 1.94 2.01 K3j 2.50 1.86 K1j 0.533 0.723 K2j 0.647 0.670 K3j 0.833 0.620 Rj 0.300 0.103 Primary and secondary orders A > B Optimum level A3 B1 Optimum combination A3 B1 C3

C 1.98 2.02 2.04 0.660 0.673 0.680 0.020 >C C3

28d UCS A B 5.51 6.67 6.38 6.36 7.10 5.96 1.837 2.223 2.127 2.120 2.367 1.987 0.530 0.236 A3 B1 A3 B1 C2

C 6.29 6.37 6.33 2.097 2.123 2.110 0.026 C2

90d UCS A B 11.61 12.92 12.33 12.37 13.37 12.02 3.870 4.307 4.110 4.123 4.457 4.007 0.587 0.300 A3 B1 A3 B1 C3

C 12.40 12.32 12.34 4.133 4.107 4.197 0.090 C3

When the factor A (geopolymer content) increases from 10% to 14%, the strength of the geopolymer stabilized soil increases significantly. Furthermore, as shown in Figs. 1 and 2, the strength of the geopolymer stabilized soil increases faster in 28d–90d than in 7d–28d. It is indicated that the strength of the geopolymer stabilized soil develops more completely in the later period, and the geopolymer content maintains a positive correlation with the strength. In addition, considering the economy of geopolymer stabilized soft soil foundation and ensuring that it meets the design requirements, the content of 14% was reasonably selected for engineering practice.

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J. Yao et al. Table 8. Range analysis of various factors on shear strength 7d Shear strength A B C K1j 0.55 0.69 0.66 K2j 0.67 0.67 0.67 K3j 0.77 0.62 0.72 K1j 0.183 0.231 0.220 K2j 0.224 0.222 0.222 K3j 0.255 0.208 0.219 Rj 0.072 0.023 0.003 Primary and secondary orders A [ B [ Optimum level A3 B1 C3 Optimum combination A3 B1 C3

The average value of UCS/MPa

Range

28d Shear strength A B C 1.30 1.88 1.68 1.76 1.76 1.71 2.11 1.52 1.77 0.433 0.628 0.590 0.586 0.588 0.571 0.703 0.506 0.561 0.270 0.122 0.029 C A3 B1 C3 A3 B1 C3

90d Shear strength A B C 2.30 3.54 3.27 3.43 3.36 3.32 4.28 3.11 3.42 0.765 1.180 1.089 1.142 1.119 1.106 1.428 1.037 1.140 0.663 0.143 0.051 A3 B1 A3 B1 C3

5 4.5 4 3.5 3 2.5 2 1.5 1 0.5 0

C3

7d 28d 90d

A1

A2

A3

B1

B2

B3

C1

C2

C3

Fact level

Fig. 1. Relationship between factors and compressive strength indexes

The Factor B (soft soil moisture content) also greatly affects the strength of the geopolymer stabilized soil, which is a secondary factor. It can be seen from Figs. 1 and 2 that with the increase of soft soil moisture content, the strength of the geopolymer stabilized soil tends to decrease, which has a negative correlation with the strength. The soft soil moisture content affects the homogeneous degree of the stabilized soil. Considering the maximum development of strength, the optimum moisture content can be obtained by referring to the experience of cement stabilized soft soil. Considering the uneven stirring of test instrument, the factor C (stirring time) has the least influence on the strength of the geopolymer stabilized soil. Therefore, according to the results of range analysis, the optimum combination is A3 B1 C3 .

The average value of shear strength/MPa

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1.6 1.4 1.2 1 7d

0.8

28d

0.6

90d

0.4 0.2 0 A1

A2

A3

B1

B2

B3

C1

C2

C3

Factor level

Fig. 2. Relationship between factors and shear strength indexes

3 Experimental Study on Mechanical Properties of Geopolymer Stabilized Soil 3.1

Experimental Study on Permeability of Geopolymer Stabilized Soil

The permeability of soil refers to the phenomenon that water penetrates through the pores of the soil, that is, the property that the pores of the soil can be transmitted by water. The permeability coefficient of soil is usually expressed by Darcy’s law. The formula is v = ki. v is the average flow velocity of the section, cm/s; i is the hydraulic gradient; k is the proportional coefficient of soil water permeability, cm/s (Table 9). Table 9. Determination data of the permeability of geopolymer stabilized soil No. 1 2 3 4 5 6 7 8 9

Mixing ratio 10 12 14 10 12 14 10 12 14

Age 7

28

90

Strength f cu =MPa 0.53 0.64 0.84 1.86 2.15 2.35 3.87 4.11 4.35

Permeability coefficient cm/s

Permeability coefficient of undisturbed soil cm/s

1:82  108 1:68  108 1:44  108 1:26  108 1:09  108 9:81  109 8:62  109 8:54  109 8:46  109

1:56  107

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Permeability coefficient of the stabilied soil×10-8

Draw the following picture. 2 1.8 1.6 1.4 1.2 7d

1 0.8

28d

0.6

90d

0.4 0.2 0 10

11

12

13

14

15

Geopolymer content (%) Fig. 3. Change diagram of permeability coefficient of the stabilized soil

In Fig. 3, it is shown that the permeability coefficient of the stabilized soil decreases with the increase of geopolymer content at the same age, which means that it is negatively correlated with the geopolymer content. At the same geopolymer content, the permeability coefficient of the stabilized soil decreases with the increase of age. When the curing age reaches 90 d, the permeability coefficient of the stabilized soil gradually decreases. It tends to be flat and a certain value. Compared with the permeability coefficient of 107 cm/s of the undisturbed soil, the stabilized soil’s can basically reach 108 cm/s or even 109 cm/s. Therefore, with the increase of strength of geopolymer stabilized soil, the structure of soil mass is gradually compact, which has significant effect on reducing its permeability. 3.2

Experimental Study on UCS and Shear Strength of Geopolymer Stabilized Soil

The shear strength of soil refers to the ultimate ability of soil to resist sliding of particles, and the compressive strength of soil refers to the ultimate strength of soil against axial pressure under unconfined conditions. Both are important indexes for characterizing the mechanical properties of polymer stabilized soils. The compressive strength and shear strength of geopolymer stabilized soil specimens at 90 d were analyzed. As shown in Table 10, the UCS of geopolymer stabilized soil has a positive correlation with the shear strength and has the same law of strength growth. When fuc = 3.75–4.73 MPa, its s = 0.753–1.523 MPa. Its compressive strength is about 3.03–5.12 times of its shear strength. When the UCS fuc = 0–4.7 MPa, the relationship between compressive strength and shear strength is approximately fcu  3:95s, that

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Table 10. Comparison of UCS and shear strength of geopolymer stabilized soil Experiment method

UCS f cu =MPa Direct shear, quick shear 3.75 3.87 3.99 4.02 4.11 4.20 4.25 4.39 4.73

Shear strength s (Mpa) 0.753 0.765 0.778 1.043 1.145 1.239 1.314 1.447 1.523

fcu =s 4.98 5.05 5.12 3.85 3.58 3.38 3.74 3.03 3.10

is,s  1=4fcu . As fcu gradually increases, the ratio of compressive strength to shear strength tends to decrease, and s  (1/5–1/3)fcu. In addition, considering the discrete characteristics of mixing piles, the standard value of shear strength is suggested to use s ¼ 0:26fcu . 3.3

Comparison of Strength Performance of Undisturbed Soils, Cement Stabilized Soils and Geopolymer Stabilized Soils

The UCS tests of undisturbed soils, cement stabilized soils and geopolymer stabilized soils at 90 d were compared under other similar conditions. The materials in the tests were undisturbed soils, DW-type geopolymer and 42:5# composite Portland cement. The contents of curing agents of both geopolymer stabilized soils and cement stabilized soils are 10%, 12%, 14%. The water content of soft soil is the natural water content kept at 49% (Table 11). Table 11. UCS test data of stabilized soils under different contents of curing agent UCS(MPa)

Content 10% 12% Undisturbed soils (US) 0.04 Cement stabilized soils (CSS) 2.98 3.22 Geopolymer stabilized soils (GSS) 3.60 4.02 fCSS =fUS 74.5 80.5 fGSS =fUS 90 100.5 fGSS =fCSS 1.21 1.25

14% 3.43 4.45 86 111.3 1.30

From the experimental data, it is shown that for stabilized soils under the same test conditions, the age is 90 d and the UCS of undisturbed soils is 0.04 MPa. When the content of curing agent is 10%, the UCS of geopolymer stabilized soils and cement stabilized soils is 3.60 MPa and 2.98 MPa respectively, which is 90 times and 75 times

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higher than that of undisturbed soils respectively. The strength of geopolymer stabilized soils is 1.21 times of that of cement soil. When the content of curing agent is 14%, the UCS of geopolymer stabilized soils and cement stabilized soils is 4.45 MPa and 3.43 MPa respectively, which is 112 times and 86 times higher than that of undisturbed soil respectively. The strength of geopolymer stabilized soil is 1.30 times higher than that of cement stabilized soil. The strength of soft soil stabilized with cement and geopolymer can be greatly improved. However, under the same content of curing agent and the same curing conditions, the strength of geopolymer stabilized soils is higher than that of cement stabilized soils. Therefore, the performance of the geopolymer stabilized soft soil is better than that of the cement stabilized soil. The geopolymer has a better effect as the content increases.

4 Conclusions In this paper, the orthogonal test method was used to analyze the influencing factors of UCS and shear strength of geopolymer stabilized soil samples. The mechanical properties of geopolymer stabilized soil were tested experimentally. The main conclusions of this paper are as follows: 1. The main factors affecting the strength of geopolymer stabilized soil are the geopolymer content and soft soil moisture content. The secondary factor is the stirring time. The optimum geopolymer content is 14%. The strength of geopolymer stabilized soil increases with the increase of curing age and geopolymer content, and decreases with the increase of moisture content of soft soil. 2. The compressive strength and shear strength of geopolymer stabilized soil specimens are positively correlated. The ratio s  (1/5–1/3)fcu was obtained. Considering the development discreteness of pile strength, it is recommended to use the standard value s ¼ 0:26fcu of shear strength. 3. The permeability coefficient of the stabilized soil can basically reach the order of 108 cm/s or even 109 cm/s. It shows that the permeability of undisturbed soil can be reduced with the reaction of geopolymer and soft soil. 4. The performance of the soft soil stabilized with the geopolymer is better than that of the cement stabilized soil. Acknowledgement. This work is supported by the National Natural Science Foundation of China under Grant No. 51608056 and 51578080, and the Transportation Department of Hunan Province, No. 2013-01. Conflicts of Interest. The authors declare no conflict of interest.

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Effect of Offset Distance on Tiered Reinforced Soil Retaining Wall Subjected to Dynamic Excitation Sudipta Sikha Saikia(&) and Arup Bhattacharjee Jorhat Engineering College, Jorhat, India [email protected], [email protected]

Abstract. The rapid growth in urbanization and demand for effective space lead to increase in application of geosynthetic reinforced soil (GRS) retaining walls in major infrastructure project like flyover etc. The researchers have conducted physical, analytical and numerical studies on performance of single tiered and multi-tiered reinforced soil wall and compared the responses. The study of multi-tiered GRS wall has not achieved the growth as the single tiered GRS wall due to its limited application. The objective of this paper is to understand the response of tiered reinforced soil retaining wall subjected to dynamic excitation. This paper emphasizes on comparative study of response of multi-tiered reinforced soil walls with single- tiered reinforced soil wall subjected to seismic excitations. A 2.8 m high finite element model of modular block facing reinforced soil wall is simulated using finite element software PLAXIS 2D. The numerical model is subjected to dynamic excitations of 0.4 g Kobe earthquake and results of the response of the numerical model are validated with shake table tests results of Ling et al. (2005). The two and three tiered walls of 9 m height with different offset distances of 0.75 m, 1.5 m and 3.0 m are simulated with validated model parameters. The construction sequence is followed in numerical model simulation and model is brought to equilibrium condition after each stage of construction. The acceleration histories of Kobe earthquake (1995) having PGA 0.4 g is applied at the base of all models. The variation of horizontal displacements, lateral pressures, maximum reinforcement loads and acceleration amplification factors of single tiered and multi-tiered walls with various offset distances are compared. It is found from the analyses that the horizontal deformation, acceleration amplification factor and maximum reinforcement load decreases with the increasing tier offset.

1 Introduction Geosynthetic reinforced soil (GRS) walls are a class of earth retaining walls where horizontal layers of metallic or geosynthetic reinforcement are placed within the wall backfill to create a composite mass comprising of soil, reinforcement layers and facing. The advantages of GRS walls including the savings in cost, ease of construction, better performance under seismic loads, design flexibility, capacity to sustain large deformations without structural distress and aesthetics made them suitable for a variety of civil engineering applications like flyovers, railways, bridge abutments, slope © Springer Nature Switzerland AG 2020 F. Tatsouka et al. (Eds.): GeoMEast 2019, SUCI, pp. 95–107, 2020. https://doi.org/10.1007/978-3-030-34242-5_9

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stability etc. The increase in height of GRS wall results in increase in tensile stresses on the reinforcements. The tensile stresses on reinforcement can be reduced by decreasing the spacing between reinforcing layers. An alternative method of reducing the tensile stresses on reinforcement is by construction of GRS walls in tiered fashion. Few guidelines are available for the design of multi-tiered walls (NCMA 1997, AASHTO 1998, FHWA 2010 etc.) and few researchers reported numerical study on multi-tiered reinforced soil walls (Leshchinsky and Han 2004, Yoo and Kim 2008, Ling et al. 2014, Mohammad et al. 2014, Bhattacharjee and Amin 2019). The present paper investigates the behavior of multi-tiered reinforced soil retaining walls under dynamic loading condition with different offset distances through numerical simulations. The study envisaged to get insight on the behavior of multitiered reinforced soil retaining wall in comparison to the single tiered reinforced soil wall.

2 Development of Numerical Model 2.1

Target Model Analysis

Ling et al. (2005) conducted large scale shaking table tests on modular block geosynthetic reinforced soil retaining wall subjected to Kobe earthquake motions. A 2.8 m high, 4 m long and 2 m wide wall constructed on a 20 cm thick soil foundation. The facing blocks were 20 cm high, 30 cm deep and 45 cm wide. The wall was backfilled with medium dense Tokachi port sand (Dr = 55%) and reinforced with PET geogrid. Geogrids length of h = 205 cm were placed at vertical intervals of 60 cm and its ultimate strength was 35 KN/m. The foundation soil has the same properties as the backfill soil but at a relative density of 90%. To prevent waves reflecting from the steel walls during shaking, 10 cm thick expanded polystyrene (EPS) boards had been placed at the front and back ends of the steel container. The geometry of the wall is shown in Fig. 1. The wall was subjected to horizontal shakings of Kobe earthquake (1995) scaled to a peak acceleration of 0.4 g.

Fig. 1. The geometry of the model used for validation (after Ling et al. 2005)

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Numerical Modeling in PLAXIS 2D

The components of shake table test as discussed by Ling et al. (2005) and corresponding finite element modelling by Liu (2014) is simulated using finite element program PLAXIS 2D. The analysis is conducted under plane strain conditions and 15 noded triangular elements are selected. The input model parameters used in PLAXIS 2D are given in Table 1. The model was fixed at the base and assumed to have roller boundaries at the sides. In the finite element model the bottom boundary is fixed in both horizontal and vertical directions while the side boundaries are fixed in only horizontal directions. To allow for absorption of stress waves, the absorbent boundary condition in PLAXIS 2D is created by placing the boundaries far apart. For dynamic excitation prescribed displacement option is used to define the seismic load. The wall is excited with maximum horizontal acceleration of 0.4 g. The acceleration time history for 0.4 g is shown in Fig. 2. Table 1. Input model parameters used in PLAXIS 2D Properties Material model Elastic modulus(kPa) Cohesion (kPa) Angle of friction(◦) Dilatancy angle w Mass density (kN/m3) Poisson’s ratio

Backfill Mohr Coulomb 156E3 1 38 8 14.30 0.33

Foundation Mohr Coulomb 156E3 1 40 8 14.30 0.33

Facing wall Geogrid Linear elastic Linear elastic 2E6

0.2

Fig. 2. Input seismic motion at peak acceleration of 0.4 g

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The geometry of the model in PLAXIS 2D is shown in Fig. 3.

Fig. 3. Deformed mesh of the validation model

2.3

Comparison of Results

Height(m)

2.3.1 Horizontal Displacement of Facing Figure 4 shows comparison between measured and predicted results of horizontal displacement of the facing wall. The maximum horizontal displacement is found to be 72 mm at the top of the wall which is similar to the measured value (70 mm) as discussed by Ling et al. (2005). The predicted and measured results show reasonably good agreement. 2.8 2.4 2 1.6 1.2 0.8 0.4 0

Ling et al. Plaxis 2D 0

20

40 60 80 100 Horizontal displacement(mm)

Fig. 4. Horizontal displacement measured by Ling et al. (2005) and FE analysis

2.3.2 Lateral Stress of Backfill The lateral stress of soil acting at the wall face is shown in the Fig. 5. The predicted and measured values of lateral stresses at the top of the wall are 2 kPa and 0.10 kPa respectively, whereas at the bottom of the wall the same are 8 kPa and 6 kPa

Fig. 5. Predicted and measured lateral stresses of backfill at the wall face

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respectively. The finite element model was able to give satisfactory agreement between the measured and predicted results. 2.3.3 Vertical Stress of Backfill The vertical stress distribution at the base of the backfill is shown in Fig. 6. The measured and predicted values show similar trends of stress distribution. The vertical stress is maximum near the facing wall where the measured value is slightly high than predicted value. The maximum predicted and measured values are respectively 240 kPa and 250 kPa respectively.

Fig. 6. Predicted and measured vertical stresses at the wall base

3 Dynamic Analyses 3.1

Numerical Modeling of Two-Tiered and Three-Tiered Reinforced Soil Wall

The preliminary design methodology of two-tiered and three tiered reinforced soil retaining wall is considered as per guidelines suggested by FHWA (2010). A 9 m high wall is selected for numerical analysis of tiered geosynthetic reinforced soil retaining walls. Single- tiered, two-tiered and three- tiered walls were considered for the analysis with three different offset distances of 0.75 m, 1.5 m and 3 m. The reinforcement lengths are calculated as per FHWA (2010) for different tiers and are shown in Table 2 below. A total of 12 number of geogrid layers are laid in all different models of tiered reinforced soil walls at a spacing of 0.6 m. The model parameters and boundary conditions considered for two and three tiered walls are same as that of the validated model. The acceleration histories of Kobe earthquake (1995) having PGA 0.4 g is applied at the base of all models. This dynamic loading is modeled by employing the prescribed displacement at the base of the foundation. Geometry of two and three tiered walls with (a) zero offset (b) 0.75 m offset (c) 1.5 m offset (d) 3 m offset distances are shown in Figs. 7 and 8 respectively.

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Position of tier Reinforcement length 0.7H = 6.3 m Upper tier 0.7H1 = 3.15 m Lower tier 0.6H = 5.4 m Three- tier (H = 9 m, H1 = H2 = 3 m) Upper tier 0.7H2 = 2.1 m Middle tier 0.7H1 = 2.1 m Lower tier 0.6H = 5.4 m

Fig. 7. Geometry of two-tiered walls with (a) zero offset (b) 0.75 m offset (c) 1.5 m offset and (d) 3 m offset distances.

3.2

Seismic Response of Tiered Walls Subjected to Kobe Earthquake

3.2.1 Horizontal Displacement of Facing The tiered walls are studied for horizontal displacement of facing and different responses are observed from each of the wall model. Figure 9 shows the comparison of horizontal displacements of the vertical wall and two-tiered walls. The wall displacement at the top of the wall is found to be 17.2 mm, 15.5 mm, 14.6 mm and 14.2 mm for offset distances of 0 m, 0.75 m, 1.5 m and 3 m respectively. Figure 10 shows the horizontal displacement of the vertical wall and three-tiered walls for different tier offset. The wall displacement at the top of the wall is found to be 17.2 mm, 15.5 mm, 12.8 mm and 10 mm for offset distances of 0 m, 0.75 m, 1.5 m and 3 m respectively The maximum deformation reduces with increasing tier offset. The graphs show that the backfill soil particle first moves towards the mid height of the walls and exerts high

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earth pressure at the mid height of the wall, near the junction of two tiers. The upper tier acts as a surcharge on the lower tier, which increases the deformation near the mid height of the wall. With increase in number of tiers, lateral facing displacement decreases for all offset at the top of wall. Due to uneven reinforcement length in the lower and upper tier, the deformation pattern is not linear.

Fig. 8. Geometry of three-tiered walls with (a) 0.75 m offset (b) 1.5 m offset (c) 3 m offset distances.

Fig. 9. Wall deformation for different offsets of two-tiered walls subjected to Kobe earthquake excitations.

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Fig. 10. Wall deformation for different offsets of three-tiered walls subjected to Kobe earthquake excitations.

3.2.2 Maximum Reinforcement Load Figure 11 shows the comparison of maximum reinforcement load for vertical wall and two-tiered walls with different offsets. The maximum reinforcement load is found to be 10.5 kN/m for vertical wall and 8.3 kN/m, 7.5 kN/m and 6.3 kN/m for offset distances of 0.75 m, 1.5 m and 3 m respectively. Figure 12 shows the comparison of maximum reinforcement load at the end of shaking for three-tiered walls with different offsets. From the graphs it can be seen that in the very bottom layers, the maximum reinforcement load in the walls are basically the same as that of the vertical wall, but the variation increased with increase in height. With the increase in number of tiers, the maximum reinforcement load in the wall with different tier-offset decreases in comparison to vertical wall.

Fig. 11. Maximum reinforcement load for different offsets of two-tiered walls subjected to Kobe earthquake excitations.

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Fig. 12. Maximum reinforcement load for different offsets of three-tiered walls subjected to Kobe earthquake excitations.

3.2.3 Acceleration Amplification The acceleration amplification ratio is defined as the acceleration at different elevations of the wall to the input peak acceleration. The acceleration amplifications at different elevations of the wall are quantified as root mean square acceleration (RMSA) amplification factor. The RMS acceleration can be calculated as: " RMS ¼

1 tZd aðtÞ2 dt td 0

#12 ðafter Kramer 1996Þ

where a(t) is acceleration time history td is duration of the acceleration record dt is the time interval of the acceleration record Figure 13 shows the acceleration amplification 10 m away from the toe for twotiered walls. For two tiered walls, amplification factors at the top of the backfill are 0.9, 2.11, 1.9 and 1.1 for offset distances of 0 m, 0.75 m, 1.5 m and 3 m respectively. Figure 14 shows the acceleration amplification of three-tiered walls at 10 m away from toe. The acceleration amplifications at the top of backfill are found to be 0.9, 2.1, 1.7 and 1 for offset distances of 0 m, 0.75 m, 1.5 m and 3 m respectively From the plot it can be inferred that even though the excitation was given at the base, the acceleration inside the backfill was minimum at the base of the wall, and the amplification factor was found to increase along the height of the wall. The acceleration amplifications are not much affected by increasing the number of tiers.

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Fig. 13. Horizontal acceleration amplifications in the backfill for two-tiered walls subjected to Kobe earthquake excitations.

Fig. 14. Horizontal acceleration amplifications in the backfill for three-tiered walls subjected to Kobe earthquake excitations.

3.2.4 Lateral Pressure Figure 15 shows the comparison of lateral soil pressure distribution exerted by the soil at the face of the wall for different tier offset of two-tiered walls. The maximum pressure is near the lower tier and reduces almost linearly with the increasing height, become minimum at the top of the wall. In tiered walls, there is a sudden increase in the lateral pressure at the junction of two tiers, which may occur due to the facing panels of the wall in the upper tier. Figure 16 shows the comparison of lateral soil pressure distribution exerted by the soil at the face of the wall for different tier offset of three tiered walls. It is observed from the analyses that three tiered walls reduce the lateral stress compared to two tiered walls considerably. The maximum lateral stress on the facing wall is not found at the base level of the backfill, but found above the base of backfill near the first layer of geogrid. This may be due to the rigid bonding between the lower most facing panels and the foundation, as considered in the analysis.

Wall height (m)

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Two-tier

10 9 8 7 6 5 4 3 2 1 0

zero offset 0.75 m offset 1.5 m offset 3 m offset

0

50

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Lateral stress(kPa)

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Fig. 15. Lateral soil pressure on the face of the wall for different tier offset of two-tiered walls subjected to Kobe earthquake excitations.

Fig. 16. Lateral soil pressure on the face of the wall for different tier offset of three-tiered walls subjected to Kobe earthquake excitations.

4 Conclusions The following section presents the conclusions drawn from the research work. • A numerical model of shake table test was developed in finite element program PLAXIS 2D and verified by comparing numerical results with physical measurements taken from shake table test on reinforced soil walls reported by Ling et al. (2005). The model was found sensitive to material properties like backfill friction and cohesion. • With increase in number of tiers, the lateral deformation reduced compared to vertical wall. Hence more the number of offsets, more stable will be the wall. With an increase in the tier-offset, the residual facing lateral displacement decreased significantly, particularly in the upper tier.

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• With the increase in number of tiers, the maximum reinforcement load in the wall with different tier-offset decreases in comparison to vertical wall. • The acceleration amplification factor was found to increase along the height of the wall and was in the range of 0.9 to 2.2 for all the walls. There is not much variation in the acceleration amplification factors with increasing number of tiers in the model walls analyzed in the study. • Tiered walls reduce the lateral stress compared to vertical wall considerably. The maximum lateral stresses are found near the bottom of the wall after which the stresses decreases almost linearly with height except at the junction of tiers where the stresses are little higher.

References AASHTO. Standard specifications for highway bridges, American Association of state highway Bridges, American Association of State highway and Transportation Officials, Washington, DC (1998) Cakir, T.: Evaluation of the effect of earthquake frequency content on seismic behaviour of cantilever retaining wall including soil structure interaction. Dyn. Earthq. Eng. 45, 96–111 (2013) FHWA. Mechanically Stabilized Earth Walls and Reinforced Soil Slopes Design and Construction Guidelines, vol. I &II. Publication no. FHWA-NHI–024, US, Department of Federal Highway Administration (2010) Hatami, K., Bathurst, R.J.: Numerical model for reinforced soil segmental walls under surcharge loading. J. Geotech. Geoenviron. Eng. 132(6), 673–684 (2006). https://doi.org/10.1061/(asce) 1090-0241(2006)132:6(673) Karpurapu, R., Bathurst, R.J.: Behaviour of geosynthetic reinforced soil retaining walls using the finite element method. Comput. Geotech. 17, 279–299 (1995). https://doi.org/10.1016/0266352x(95)99214-c Leshchinsky, D., Han, J.: Geosynthetic reinforced multitiered walls. J. Geotech. Geoenviron. Eng. 130(12) (2004). https://doi.org/10.1061/(asce)1090-0241(2004)130:12(1225) Ling, H.I., Cardany, C.P., Sun, L.-X., Hashimoto, H.: Finite element study of a geosyntheticreinforced soil retaining wall with concrete-block facing. Geosynthetics Int. 7(3), 163–188 (2000) Ling, H.I., Leshchinsky, D.: Parametric studies of the behavior of segmental block reinforced soil retaining walls. Geosynthetics Int. 10(3), 77–94 (2003). https://doi.org/10.1680/gein.10.3.77. 37074 Liu, H.: Long-term lateral displacement of geosynthetic-reinforced soil segmental retaining walls. Geotext. Geomembr. 32(2012), 18–27 (2011) NCMA. Segmental Retaining Walls Seismic Design Manual. National Concrete Masonry Association. Bathurst, R.J. (ed.) 1st edn. Herndon, Virginia, USA (1998) Reference Manual: PLAXIS 2D-Version8; Brinkgreve, R.B.J. (ed.) Delft University of Technology & PLAXIS; The Netherlands Richards, R., Elms, D.G.: Seismic behavior of gravity retaining walls. J. Geotech. Eng. Div. ASCE 105(GT4), 449–464 (1979) Rowe, R.K., Ho, S.K.: Continuous panel reinforced soil walls on rigid foundations. J. Geotech. Geoenviron. Eng. 123(10), 912–920 (1997). https://doi.org/10.1061/(ASCE)1090-0241 (1997)123:10(912)

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Scientific Manual: PLAXIS 2D-Version8; Brinkgreve, R.B.J. (ed.) Delft University of Technology & PLAXIS; The Netherlands Seed, H.B., Whitman, R.V.: Design of earth retaining structures for dynamic loads. In: ASCE Specialty Conference: Lateral Stresses in the Ground and Design of Earth Retaining Structures, pp. 103–147 (1970) Tatsuoka, F.: Geosynthetic-reinforced soil retaining walls as permanent structures. In: Proceedings of Indian Geotechnical Conference 2002, Allahabad, India, pp. 681–699 (2002) Tutorial Manual: PLAXIS 2D-Version8; Brinkgreve, R.B.J. (ed.) Delft University of Technology & PLAXIS; The Netherlands

Review of Process Control and Assurance for Optimized Seaming Condition Optimization of Woven Geotextiles to Improve Stability in Soft Soil Structure Han-Yong Jeon(&) Department of Chemical Engineering, Inha University, Incheon, South Korea [email protected]

Abstract. In this study, nonwoven and woven geotextiles were used and the different seaming methods were applied to review the seam properties through comparison of tensile strength retention by seaming methods. Also, bonding agents were applied to strengthen the seaming effect on the seamed part and reduction factors of woven geotextiles were analyzed by field installation conditions. Finally, process control and assurance for optimized seaming condition of woven geotextiles were reviewed to improve stability in soft soil structure.

1 Introduction The woven geotextile is a geotextile product woven in the direction of the warp yarn using weaving technology and has been widely applied to civil works such as separation, protection and reinforcement. In Korea, it is widely used for reinforcement of soft ground and reinforcement of road foundation. [1, 2] The soft ground is widely distributed on the west coast and the south coast. Woven geotextile is essential for port facilities and coastal complex development. Woven geotextile is installed by sewing in the field suitable for the construction area. However, in order to secure the equipment running ability for entering the equipment when improving the landfill of soft ground, Woven geotextile damage can cause accidents such as personal injury and equipment evacuation. Therefore, as the use of woven geotextiles increases, it is necessary to secure more than 50% of the strength of woven geotextile base materials that have not been sewn to the strength of joints by the sewing method in order to secure stability by entering heavy equipment into soft ground. [4, 5] Currently, woven geotextiles are used to seal the width and width of the geotextile in order to secure the workability when applied to the field, and 2, 3 or 6 seam lines are given in consideration of each construction situation. Generally, since the sewing efficiency by the sewing line is inconsistent, the method of applying the sewing line is appropriately selected according to the construction conditions. On the other hand, when the external load is applied to the woven geotextile stitching area, it is common that the stitching area is destroyed due to the stress concentration phenomenon occurring in the stitching line. The destruction of these seams is the most serious problem of seam strength degradation, which has a great influence on the stability of the entire woven geotextile style, and the © Springer Nature Switzerland AG 2020 F. Tatsouka et al. (Eds.): GeoMEast 2019, SUCI, pp. 108–120, 2020. https://doi.org/10.1007/978-3-030-34242-5_10

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reinforcing effect of the soft ground is drastically reduced. Therefore, a reasonable method of strengthening the woven geotextile seamed part at the site construction should be proposed, and the reduction factor and thus the long-term performance of the woven geotextile seam strength reduction should be considered. A reasonable method of strengthening the woven-geotextile-style seamed part at the site construction should be proposed, and the reduction factor and thus the long-term performance of the woven geotextile seam strength reduction should be considered.

2 Experimental 2.1

Preparation of Seamed Geotextiles by Adhesive Reinforcing

Woven geotextile (; of plain weave and design strength, 6 ton/m) was used by weaving with 1,600 denier polyester yarn, which is most applied in domestic soft ground reinforcement. As a result of reviewing the number of seamed lines in order to improve the seaming efficiency of the seamed area, the woven geotextile with 6-line seamed part by adhesive were shown in Fig. 1. The used adhesives are shown in Table 1. As the adhesive, cyanoacrylate and epoxy adhesives were selected as suitable adhesives referring to adhesive component and solidification time, in order to strengthen the sealing area, securing the seal ability, and efficiency and time of the sealing process. Adhesive agent was added to the seam of the woven geotextile adhesive area, and the seam strength was measured and analyzed according to ISO 10320 and 6-line seamed samples without adhesive was used as a reference sample for comparison.

Fig. 1. Photograph of reinforced 6-line seamed part of woven geotextile by adhesive agent.

Table 1. Adhesive agent for reinforced seaming of woven geotextile Adhesive Agent

Specification Cyanoacrylate type CA-1 CA-2 Epoxy type EP-1 EP-2

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Here, the cyanoacrylate-based adhesive is an ultrahigh-speed adhesive known as an instant adhesive, which is made of a cyanoacrylate monomer (methyl, ethyl, etc.) and is a single component product without a solvent. Polymerization is due to moisture on the surface of the material, and can instantly cure almost any surface or material, especially with very thin bonding lines. Strong adhesion to all materials, impact resistance, heat resistance, moisture resistance, odorless, non-whitening adhesive work quickly and safely. CA-1 (AXIA, Henkel Co., USA) and CA-2 (Henkel LOCTITE®401™, Henkel Co., USA) are cyanoacrylate type instant adhesives. They are applied with temperature, primer, and viscosity, initial and complete curing time, Glass transition temperature, and storage period. On the other hand, the epoxy adhesive is a thermosetting resin adhesive based on an epoxy resin, a curing agent, a filler, a diluent, and other additives, and has excellent mechanical properties after curing, strong adhesive force, excellent heat resistance and electrical insulating properties, There is no by-product in the reaction, and it is widely used in all industrial fields such as home, industrial, construction, electricity, automobile, and aircraft. (DEVCON, ITW Performance Polymers®, USA) and EP-2 (EP-1), which are widely used when durability is needed, (3M Scotch Weld DP100, 3M (Minnesota Mining & Manufacturing, USA), which means two liquid types and one liquid type, respectively. 2.2

Assessments of Engineering Performance of Seamed Geotextiles

Installation Damage In accordance with ISO 10722, installation damage was assessed during construction of the woven geotextile stitching area, and then the strength retention ratio (%) was measured and analyzed according to ISO 10320. In order to compare the seam strength, the endurance test of the 6-line seamed without adhesive was also conducted. Chemical Resistance The tensile strength retention ratio by chemical resistance was measured and analyzed according to US EPA 9090 Test Method and ASTM D5322 under the conditions of constant temperature and pH value after immersing for 60 days at 23 °C, 50 °C and pH 5, 11, 13, respectively. For the comparison of the seam strength, the chemical resistance test of 6-line seamed without adhesive was also performed. Salt Water Resistance The samples were immersed for 30 days at room temperature in a NaCl 3.8% solution at the concentration condition simulating seawater conditions, and then the strength retention (%) was measured and analyzed according to ISO 10320. In order to compare the seam strength, the salt water resistance test of the 6-line seamed without adhesive was also conducted. Frictional Property In order to evaluate the frictional property of soil with soft soil, shear characteristics of the seams were tested and analyzed according to ISO 12957-1. In order to compare the seam strength, the frictional properties of the 6-line seamed without adhesive were also tested.

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Analysis of Long-term Seam Strength The reduction factors affecting the seam strength of the woven geotextile were determined and then the long-term performance of the seamed was analyzed and evaluated.

3 Results and Discussion 3.1

Reinforced Seam Strength by Adhesive

Figure 2 shows the seam strength according to adhesive type. It was found that the seam strength was increased when the adhesive was added, and the seam strength was significantly improved when the cyanoacrylate adhesive was added to the adhesive rather than the epoxy adhesive. In the case of the epoxy adhesive, it is considered that the strength deterioration occurred due to the embrittlement of the polyester fiber caused by the curing agent during the pot life. In the case of the cyanoacrylate adhesive, the viscosity and the adhesive strength were larger than those of the epoxy adhesive. In the case of the cyanoacrylate adhesive, CA-2 shows better seaming effect than CA-1 and this is due to higher viscosity and adhesive strength than CA-1.

Fig. 2. Seam strength of reinforced 6-line seamed part of woven geotextile by adhesive agent.

3.2

Installation Damage of Reinforced Seamed Part by Adhesive

Figure 3 shows the percent retention (%) of the seam strength after testing for damage to the woven geotextile seams in accordance with ISO 10722. As in the case of the adhesive strength of Fig. 3, it can be seen that when the adhesive is added, the damage of the sealing portion is less than when the adhesive is not added. In the case of the cyanoacrylate adhesive, it was found that there was less damage to the seamed part. In addition, the cyanoacrylate-based adhesives were found to have a relatively low seamed damage to CA-2.

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Fig. 3. Seam strength of reinforced 6-line seamed part of woven geotextile by adhesive agent for installation damage test.

3.3

Chemical Resistance of Reinforced Seamed Part by Adhesive

Figure 4 shows the retention strengths of the woven geotextile seams immersed for 30 days at 23 °C and 50 °C and pH 5, 11, 13, respectively. It was found that when the adhesive agent was added, the chemical resistance of the seamed region was better than that of the case where the adhesive agent was not added, and in the case of the cyanoacrylate adhesive agent addition, the chemical resistance was superior to that of the epoxy adhesive agent. In the case of the cyanoacrylate adhesive, it was found that the chemical resistance of the sealing portion of CA-2 was excellent.

Fig. 4. Seam strength retention of reinforced 6-line seamed part of woven geotextile by adhesive agents for chemical resistance test

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Salt Water Resistance of Reinforced Seamed Part by Adhesive

Figure 5 shows the rate of change in the seam strength of the woven geotextile dipped in a solution of NaCl 3.8% at room temperature for 30 days. As shown in Table 2, when the adhesive was added, the salt resistance of the seamed area was better than that of the adhesive without the adhesive. It was found that the cyanoacrylate-based adhesive agent was more excellent in the brine resistance than the epoxy adhesive agent, and that the cyanoacrylate-based adhesive agent was excellent in the brine resistance of the sealing portion of CA-2.

Fig. 5. Seam strength retention of reinforced 6-line seamed part of woven geotextile by adhesive agents for simulated sea water resistance test

3.5

Frictional Properties of Reinforced Seamed Part by Adhesive

Figures 6, 7 and 8 shows the shear properties of the woven geotextile stitched area measured in accordance with ISO 12957, and it was confirmed that the frictional properties of the stitched area were excellent in the case of the cyanoacrylate-based adhesive. However, in the case of the epoxy adhesive, it was found that the frictional properties were lowered compared to the samples without the adhesive, which is presumed to result from the embrittlement of the polyester fibers caused by the curing agent during the service life. In addition, as shown in the above-mentioned experiment results, it is believed that the cyanoacrylate-based adhesive has a higher viscosity and higher adhesive strength than the epoxy-based adhesive. This suggests that the frictional properties with the soil are affected by the seam strength when the soft ground is applied, and the frictional properties are improved when the seam strength is improved by the adhesive. In the case of cyanoacrylate adhesives, CA-2 shows excellent stiffening effect because the stiffening effect of the seamed is larger than CA-1.

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(a) Direction to seaming line, 90° (A) and 180° (B)

(b) Under normal stress 30kPa (A), 50kPa (B), 70kPa (C) Fig. 6. Shear frictional properties of reinforced 6-line seamed part of woven geotextile by adhesive agents

3.6

Reduction Factors on Reinforced Seamed Part by Adhesive

In each of the above test conditions, the reduction factors affecting the seam strength of the woven geotextile were determined and then the total reduction coefficient was obtained. When the woven geotextile is applied to the soft ground, the maximum tensile strength obtained from the index test is not applied as it is, but the correction factor considering the reduction factor of the tensile strength considered in the construction condition should be applied. At this time, the correction factor is referred to as a reduction factor, and the type of the reduction factor to be considered is selected and determined by the application field. The seam strength shall also be considered in the same dimension as the tensile strength and shall be calculated by applying the corresponding reduction factor to the following formula which determines the long-term permissible strength.

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(a) Maximum shear stress

(b) Shear frictional coefficient

(c) Frictional angle Fig. 7. Shear frictional properties of reinforced 6-line seamed part of woven geotextile in direction to seaming line, 90° by adhesive agents under normal stress 30 kPa (A), 50 kPa (B), 70 kPa (C)

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(a) Maximum shear stress

(b) Shear frictional coefficient

(c) Frictional angle Fig. 8. Shear frictional properties of reinforced 6-line seamed part of woven geotextile in direction to seaming line, 180° by adhesive agents under normal stress 30 kPa (A), 50 kPa (B), 70 kPa (C)

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 Tallow ¼ Tult

 ¼ Tult

1 ðPRF Þ

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 ð1Þ



1

ðRFID  RFCR  RFCD  RFBD  RFJC . . .Þ

;

ð2Þ

(where Tult = ultimate tensile strength, Tallow = allowable tensile strength, RFID = reduction factor by installation damage, RFCR = reduction factor by creep deformation, RFCD = reduction factor by chemical degradation, RFBD = reduction factor by biological degradation, RFJC = reduction factor by junction property etc.). In other words, instead of the reduction factor considered in the above equation, the reduction factor that reduces the permissible seam strength, the long-term performance of the seamed, must be considered. Considering that the location of the soft ground is a coastal area instead, the long-term permissible strength is calculated by the reduction coefficient as shown in the following formula by applying the reduction coefficient by RFSS = salt water resistance considering the simulated seawater solution.  Tallow ¼ Tult

1 ðPRF Þ



 Tult

 1 ; ðRFID  RFCD  RFSS Þ

ð3Þ ð4Þ

(where Tult = ultimate tensile strength, Tallow = allowable tensile strength, RFID = reduction factor by installation damage, RFCD = reduction factor by chemical resistance, RFss = reduction factor by salt water resistance etc.). Figures 9 and 10 shows the reduction factor and the total reduction factor, respectively, obtained from the changes in the engineering properties. 3.7

Long-Term Permissible Seam Strength of Reinforced Seamed Part by Adhesive

Figure 11 shows the long-term permissible seam strength obtained by considering the range of corresponding reduction factors affecting woven geotextile-style seam strength. Here, the long-term permissible seam strength was obtained without considering the creep deformation or the reduction coefficient due to hydraulic behavior. In the case of CA-2, which is the best cyanoacrylate adhesive in the present study, it was confirmed that the retention strength was higher than 94%.

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(a) For installation damage, RFID

(b) For chemical resistance, RFCD

(c) For salt water resistance, RFSS Fig. 9. Reduction factor of reinforced 6-line seamed part of woven geotextile by adhesive agents

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Fig. 10. Total reduction factor of reinforced 6-line seamed part of woven geotextile by adhesive agents

(a) Total reduction factor, RFTOTAL

(a) Long-term seam strength Fig. 11. Total reduction factor and long-term seam strength of reinforced 6-line seamed part of woven geotextile by adhesive agents

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4 Conclusion Different reinforcement effect was confirmed according to the type of adhesive, and it was found that the retention strength of the seam strength at the time of bonding was excellent in the case of the cyanoacrylate adhesive rather than the epoxy adhesive. In addition, CA-2 among the cyanoacrylate adhesives showed a retention ratio of 94% or more, and the reduction factors affecting the seam strength at the time of field application directly affect the tensile strength and the sealing strength of the woven geotextile. It is believed that the engineering properties, total reduction factors and longterm performance of the adhesives used contributed to improving the seam strength of the woven geotextile. Therefore, we believe that the woven geotextile seamed area should be reinforced because it has a significant effect on the stability of the geotechnical structure. Finally, this study suggests effective reinforcement method of the seamed part, and it is expected that it will play a big role in improving the workability and stability if improvement of the seamed equipment to improve the workability of the soft ground in the field seaming is progressed. Acknowledgments. This research was supported by a grant (18RDRPB076564-05) from Regional Development Research Program funded by Ministry of Land, Infrastructure and Transport of Korean government.

References 1. FHWA: Geotextile Design & Construction Guidelines, U.S. Dept. of Transportation Federal Highway Administration, Publication No. FHWA HI-90-001, pp. 24–46 (1989) 2. GRI: GRI Standard Test Methods on Geosynthetics. Drexel University, Philadelphia (2007) 3. Holtzs, R.D., et.al.: Geosynthetic Design and Construction Guidelines. U.S. Dept. of Transportation Federal Highway Administration, Publication No. FHWA HI-95-038, pp. 27– 105 (1995) 4. IGS Educational Resource: Geosynthetics Functions, International Geosynthetics Society (2010) 5. Jewell, R.A.: Soil Reinforcement with Geotextiles. CIRLA Special Publication, Thomas Telford, Westminster (1996). Chapter 5 6. Koerner, R.M.: Designing with Geosynthetics, 6th edn. Xlibris Co., Bloomington (2012) 7. Leu, W., Luane T.: Applications of geotextiles, geogrids, and geocells in Northern Minnesota. In: Geosynthetics Conference 2001, Portland, Oregon, USA, pp. 809–821 (2001) 8. Rollins, A.L., et.al.: Evaluation of field seams quality by the impact test procedure. In: Geosynthetics 1991, February, Atlanta, GA, USA, pp. 223–237 (1991)

Celebrating Reinforced Soil Structures A Historic Review from the Mid-60’s Original Concept to Today’s Design and GOOD Construction Practice Using Site-Won and Other Non-‘Standard’ Reinforced Soil Fills Chaido Doulala-Rigby(&) Tensar International Limited, Blackburn, UK [email protected] Abstract. The first use of High Density Polyethylene (HDPE) polymeric geogrid reinforcement in civil engineering was to reinforce and construct a 2.5 m high temporary reinforced soil wall at Newmarket/Silkstone colliery in West Yorkshire, UK in 1980, just 2 years after the first polymeric geogrid was invented by Dr Mercer in 1978 in Blackburn, UK. HDPE geogrid Reinforced Soil Retaining Wall (RSRW) Systems have since been widely used around the world forming various geometries, reaching unprecedented retaining heights in excess of 60 m and serving various functions from supporting open air golf courses to airport runways. This keynote will give a historic insight on how HDPE geogrid RSRW Systems have evolved in the past 40 years through presenting 10 different case studies, showcasing different types of non-standard reinforced fills including both site-won and purpose-made reinforced fill, the challenges they presented and the lessons learned. By describing the use of variable, non-standard reinforced fills ranging from site-won cohesive fill, to site-won chalk, to site-won mine stone waste, to locally sourced waste pulverized fuel ash, to landfill waste site-won fill, and others, it will showcase the selection criteria and applicability of these various fills depending on the performance requirements of the end structure. It will also highlight critical issues that need to be taken into consideration when using non-standard reinforced soil fills, both at design stage and during construction, such as bespoke site testing as well as contingency and remediation plans to cater for inclined weather or for when site testing does not meet the required performance. The ultimate purpose of this Keynote is to, as the title suggests, celebrate polymeric geogrid reinforced structures and manifest how they have become established as reliable alternatives to conventional reinforced concrete structures. In many situations, the discovery of polymeric geogrids has opened up possibilities for the construction of extraordinary retaining structures that would not otherwise be feasible or would be extortionately expensive, like the 60 m high polymeric reinforced soil walls featured as alternatives to conventional concrete viaducts in Fujairah, UAE, thus allowing rapid construction and providing earth retaining solutions resulting in attractive, stable, cost effective and maintenance free structures for their 120 years design life.

© Springer Nature Switzerland AG 2020 F. Tatsouka et al. (Eds.): GeoMEast 2019, SUCI, pp. 121–150, 2020. https://doi.org/10.1007/978-3-030-34242-5_11

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C. Doulala-Rigby Keywords: Case histories Resilience
Sustainability


Geogrids
Reinforced soil
Retaining walls


1 Historic Review The concept of Earth or Soil Reinforcement in not new. As with many fundamental engineering concepts, the very origin of them can be found in nature; in the case of ‘reinforced soil’ the very obvious examples are that of animal and birds nest building with twigs and clay and in the action of tree roots. The earliest man-made surviving examples of the same (in principle) technique date thousands of years back and are perfectly demonstrated in structures that have passed the test of time and are still standing, such as the construction of Ziggurats in ancient Mesopotamia over 3000 years ago using layers of weaved reeds laid horizontally on layers of sand and gravel (Bagir et al. 1944), the Great Wall of China during the 7th Century BC using mixtures of clay and gravel reinforced with tamarisk branches (Jones 1985) and numerous Roman applications using reed or wood reinforcement. Such notable Roman structures include the reed-reinforced earth levees along the river Tiber and the first-century Roman Army timber wharf for the Port of Londinium, that was discovered in London, UK preserved in the Thames mud for 1200 years, believed to have been 1.5 km in length, 2 m in height and formed from oak baulks up to 9 m in length embedded in the back fill (Bassett et al. 1981). The fundamental engineering concept of soil reinforcement, as originally conceived thousands of years ago, was very simple and was based on the improvement of the soil behavior by placing organic ‘geo-elements’, in between the soil layers to add strength, thereby creating a composite ‘geosystem’. The very essence of reinforced soil fundamental principles remains the same to date. The only difference is that the ‘geo-components’ that comprise modern ‘geosystems’ now a days, in most developed places around the world, are no longer organic but have been replaced by new man-made ‘geo-synthetic’ materials. Having said that, organic reinforcement is still used in places like Africa to construct ‘adobe’ bricks out of clay and hay to build their dwellings, an ancient technique that can be traced back in Bible readings (book of Exodus, Ch. 5). In more recent history and looking back at the early 1800’s, examples of soil reinforcing techniques can be found in military engineering, although very little was published. Sir Col Charles Pasley, a visionary and a pioneer military engineer, through carrying out a number of comprehensive trials reinforcing the backfill of retaining walls with brushwood, wooden planks or canvas, he proved that the inclusion of such layers resulted in significant reduction to the lateral earth pressures (Pasley 1822). The ‘modern’ concept of reinforced soil saw significant development over the last century, with a number of inquisitive engineers experimenting with various soil reinforcing elements like Munster’s wooden reinforcing arrays (Munster 1930), Coyne’s steel tie members with small end anchors (1930’s) and Casagrande’s high-strength membranes (Westergaard 1938), to mention but a few of some of the most notable ones. But the man that is known to have scientifically captured the mechanism and can be identified as officially establishing the concept of modern “Reinforced Earth”, was

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Henri Vidal who patented the use of steel strips to reinforce soil and coined the term Reinforced Earth ‘Reinforced Earth®’ in 1963. Vidal’s only valid patent related to the use of metal strip reinforcement with protrusions no greater than 3 mm. Earlier in 1959 another French engineer, Lallemand, had patented the use of metal strips with protrusions greater then 3 mm. Importantly any reinforcement in the form of a grid or anchor was not covered by Vidal’s patent. The term ‘Reinforced Earth®’ is a trademark in some countries. During the 1970s, UK, started to use the term ‘Reinforced Soil’ to keep the peace (Jones personal communication 2019). Around the same time and following along the same fundamentals but further developing the mechanism of ‘true’ soil reinforcement and experimenting with a pioneer manufacturing process, Dr Brain Mercer, a British inventor and entrepreneur, introduced a brand new polymer product, which comprised a continuous monolithic grid of polymer ribs and transverse bars with integral junctions and regular apertures. This new product was named ‘geogrid’, a term that was introduced by Prof Peter Wroth circa 1981/2 (Jones personal communication 2019) to describe this new product; Dr Mercer patented this new product as Tensar® geogrid in 1978 in Blackburn, UK. The fundamental mechanism enhancement and justification of the term ‘true’ soil reinforcement, in the author’s opinion, is attributable to the additional reinforcing effect due to the interlock of the soil particles within the rigid geogrid apertures, a mechanism that cannot be developed in soil reinforced with steel or polymer strips or ladders (Fig. 1).

Fig. 1. Newmarket/Silkstone reinforced soil retaining wall

We should not underestimate the immense achievement of introducing a new technology and a new material into civil and structural engineering over such a short time scale, especially at an era when no design standards included such construction materials yet. Most long-term engineered structures are required to be designed for a life of 120 years and predicting the acceptable lives for materials whose creep and fracture resistance were largely unknown from relatively short-term tests presented formidable problems. Acceptance of innovations in the construction sector also

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depends crucially on the adoption of authoritative design guidance and this is no simple matter. At the time, there were no design standards that covered the use of polymer geogrids to reinforce soil. The closest national standards available was BE 3/78, that allowed the use of any material which could meet the specification with regard to strength and durability but with no further specific guidance. At the time, West Yorkshire Metropolitan County Council (WYMCC) was the only authority in the UK with experience in the design and construction of reinforced soil. Pilkington Fibretain glass reinforcement was used on M62 in 1973 and was the standard material use by WYMCC for a number of years. Paraweb reinforcement developed by ICI was also used in Yorkshire both by WYMCC and the National Coal Board. WYMCC also constructed structures in the 1970s using steel grid reinforcement. Tensar’s John Templeman, a Civil Engineer that worked with Dr Mercer at the time, reached out to WYMCC in 1978/9 to ask for their help in developing the application of these new continuous monolithic polymer geogrids. Colin Jones, who work for WYMCC at the time, was the first civil engineer that had the vision, innovative spirit and dared to compile his sound engineering judgement with the first short term strength test results of these new polymer products to produce the very first high density polyethylene (HDPE) geogrid reinforcement design to construct a temporary 2.5 m high reinforced soil wall at Newmarket/Silkstone colliery in West Yorkshire, UK in 1980, using locally sourced mine waste (Jones and Doulala-Rigby 2014) and to reinforce site-won soil blast bunds for the Ministry of Defense to protect various ammunition storage locations in the UK and overseas (Jones personal communication & UK Corps of Engineers personal communication 2019). In the past 40 years since their invention, a plethora of design standards and codes were formulated to capture the calculus design mechanism of the interaction between soil and polymer geogrid reinforcement. The first British design code published specific to Reinforced Soil design and specifically including ‘extensible’, i.e. polymer reinforcement, was the British Standard BS 8006 leading the way in 1995 for the first time, after it had been in preparation for 11 years since 1984, when the first British Standards committee on reinforced soil was formed. The first American Standards followed with the publication of “Guidelines for Design, Specification, and Contracting of Geosynthetic Mechanically Stabilized Earth Slopes on Firm Foundations” by the Federal Highway Administration (FHWA) in 1993 and then other countries like Germany, Holland, Australia, Hong Kong, South Africa and others followed with their own design standards publications. The first Symposium on Polymeric Geogrids ever to be organized took place in 1984 under the auspices of the Institution of Civil Engineers in the UK. The Symposium encouraged the wider applications of polymer geogrids. It also stimulated the application of many of the research findings and encouraged on-going investigations. Since then, much has been learnt about the engineering properties of the pioneer HDPE monolithic polymer geogrids and many other polymer geogrids of varying consistency and structure have come into the market by various manufacturers, each functioning in their own unique way. Polymer geogrid reinforced soil technology has become a trusted construction method and has been used in the construction of thousands of ground separation structures supporting major railways, trunk roads and motorways as well as forming marine walls, airports, bridge abutments or simply supporting open area recreation areas.

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This paper, through 10 case studies from all over the world, will present the versatile nature of the structures that can be materialised depending on the type of reinforced fill available while focusing on the main considerations that need to be addressed when using non-standard reinforced fills, the challenges they presented and some lessons learned. The order of presentation of the case studies in this paper does not follow a strict chronological sequence but it starts by showcasing extreme structures that can be achieved when the best of reinforced fills is available, moving on to equally impressive but less extreme structures with fills of lesser quality and onto structures formed from unusual, waste or not so well known fills.

2 Important Considerations When Using Non-standard Fill Regardless of the type of reinforced fill available or planned to be used for the construction of a reinforced soil structure, the first vital activity that needs to be carefully handled at the onset of every contract is a clear, un-ambiguous and honest conversation between all stake holders, and in particular between the reinforced soil designer and the constructor who will build it. As project procurement is becoming more and more fragmented in recent years, the risk of creating gaps in responsibility is becoming higher and indeed critical. The modern tendency is towards the ‘specialised’ components of a structure, and reinforced soil is often viewed as such, being procured on a design-and-construct basis, which can lead to gaps in responsibility that would not otherwise arise if the design and supervision responsibility resided with a single entity. It also creates a situation where the requirements of the ‘specialised’ designer may conflict with the priorities and interests of the main contractor. It is therefore vital that all parties understand the scope and importance of the various components specified in a project, both ‘specialised’ or ‘standard’, to avoid, for example, unauthorized material replacements or indeed omissions that could lead to catastrophic consequences. A typical example is the unauthorized replacement of a high spec reinforced fill with one of lesser quality due to the lack of understanding of the importance of the fill quality in a reinforced soil structure performance. Another typical example is the responsibility for the assessment of global stability of a reinforced soil structure often falling between the overall project designer and the designer/supplier of the proprietary reinforced soil retaining structure supplier, which if not clearly defined at the beginning of the project can lead to painful delays and indeed project failures. And another typical example is that, often, the responsibility of designing and coordinating temporary works vital to support the excavated retained fill prior to the installation of the reinforced soil structure is overlooked, becoming an oversight which could lead to a potentially major collapse, possibly with catastrophic consequences too. So an allstake-holder ‘kick-off’ project meeting to make sure everybody understand the scheme and the importance of its components is vital to the successful execution of any project. Some other, more practical but equally important considerations, specific to when using ‘non-standard’ fill are listed below, as derived from industry and from my personal *25 years’ global experience serving as a Geotechnical Engineer:

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• Non-standard, site-won fill, can be highly variable, as often is the case, so appropriate mixing and adequate number and type of soil samples must be taken for laboratory testing in order to be representative. • Non-standard fill needs to be appropriately classified and the use of appropriate testing needs to be assigned to derive its required design properties. The most usual laboratory testing to derive shear strength parameters to be used in the design of any reinforced soil structure is a shear box test; use of appropriate size shear box is vital and in case where stones are present in the site won fill, careful consideration of the amount of stone removal must take place together with the effect of such stone removal to the behaviour of the soil. If too many stones are removed, then the derived shear strength properties will not be representative as the ‘stone-less’ soil samples are likely to behave in a different manner to the actual site-won fill from the available raw stockpile. • Appropriate screening of site won material and removal of extra-large boulders >125 mm must take place preferably prior to construction commencement or during construction, if necessary; case study 3.3 below refers. • Unlike the reasonably fast shearing rate in the order of 0.8 mm/min that is expected for coarser gravelly sandy fills, if finer/clayey rich materials are proposed as reinforced fill, the rate of shearing in the shear box testing needs to be carefully specified. For a fine sandy clay, for example, a very low shearing rate in the order of 0.001–0.009 mm/min is expected to be used, after the specimens are submerged/saturated and then consolidated, in which case testing would take at least 6 days! This is the only correct way for slow draining clay fills to be tested in order to produce true, representative ‘drained’ shear strength parameters to be used in design rather than assigning the ‘typical’ fast rate of shearing, which would lead to false, non-representative shear strength design properties and consequently to unsafe design. • Depending on the specified reinforced fill material nature, whether imported, sitewon or man-made, especially if it is particularly angular, unusually fine, rounded, or in general ‘non-standard’, and has not been used with the specific geogrid before, bespoke, project specific laboratory shear and pull out testing, and maybe even damage or abrasion testing, will need to take place preferably prior to design commencement but definitely prior to construction, to investigate the affect it will have on the geogrid that will be used to reinforce, as well as to derive representative interaction parameters between geogrid and fill, such as pull out and sliding factors to be used in the design. • Careful chemical analysis of ‘non-standard’ fill, especially when elevated pH levels are suspected must be carried out. HDPE is largely inert to chemical attack and to environments with pH4–pH12.5 but not all soil reinforcement geogrids are, so added FoS should be incorporated in the design, as required, or avoidance of specific polymer geogrids all together if seemed overly susceptible, based on the findings of the chemical analysis. • Likewise, in projects where the ‘non-standard’ fill comprises recycled waste fill rich in contaminants like bisphenol A (BPA) or polychlorinated biphenyl (PCB) or/and similar apart from added personal protection equipment (PPE) considerations, special consideration must also be given on the design of the capping layer over the

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•

•

•

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reinforced soil structure to effectively contain the contaminants and prevent future contamination through leaching or even through airborne paths – case study f will elaborate further on this. For recycled site-won fills, especially fills that are derived from demolition waste rich in brick fragments for example, additional abrasion testing maybe necessary to establish crushability under compaction so that appropriate plant and method of compaction can be used that won’t lead to fill particles breaking down under compaction. Similar considerations should be taken into account when other ‘non-standard’ fills, such as light weight aggregate (LWA), is used as reinforced fill, in which case close co-operation with the LWA provider is imperative and crucial so that their advice for handling and compacting LWA is strictly followed; case study f will elaborate further on this. In-situ specific testing and site controls should be prepared, such as bespoke, sitespecific schedule of contingency measures to be followed on site for projects where highly cohesive soil is used and challenging climatic conditions are expected – case study z below elaborates further on this very important issue. DRAINAGE, DRAINAGE, DRAINAGE! Due consideration must always be given to drainage while using any reinforced or conventional fill but drainage becomes even more important for finer fills, or moisture sensitive fills like chalk and/or Pulverised Fuel Ash for which, specific measures need to be implemented during construction and are also given in various national design standards – case studies x & y refer below.

3 Case Studies 3.1

Dubai Fujairah Freeway Walls, 2012

In early 2006 and in order to facilitate the development of the region, the Ministry of Public Works of United Arab Emirates, decided to construct a 4-lane motorway (10 lanes in total, including 2 hard shoulders), which was to link Fujairah with Dubai. The 10-lane road was aligned to navigate through mountainous landscape. Intense rainfall which occurs typically once a year has formed deep gullies and valleys within the terrain. The new road had to bridge over the existing valleys and cut through existing severe gradients of the mountains. The original conforming design was the construction of major viaducts across the valleys. However, the construction of viaducts would have been incredibly costly and challenging due to the lack of access, water resources and difficulties of facilitating concrete curing in the extreme temperatures and the arid conditions. Coupled with the large amounts of site-won good quality virgin Gabbro available due to blasting to create access of the new freeway through the arid mountains, HDPE geogrid reinforced soil retaining wall (RSRW) technology was opted for the construction of 29no individual reinforced soil retaining walls for the project. 25no of them form single tier walls with maximum heights up to 22 m; four of them formed major tiered embankments made of two or three tiers with maximum heights up to 60 m with large culverts incorporated at the base of the embankments to

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deal with the excessive, expected storm water flow in the wadis during the infrequent but severe rainstorms. Figure 2 below shows a typical elevation of the 60 m high 3-tier walls with the culvert at the base. The total reinforced soil wall constructed was in excess of 100,000 m2.

Fig. 2. Typical elevation of the 3-tier, 60 m high Fujairah Walls

The RSRW system adopted comprise four major components, namely, the concrete modular face block, the HDPE uniaxial geogrid, the polymeric mechanical block connector and the reinforced fill material, all dry laid. The concrete modular face blocks that are 200 mm wide, 200 mm deep, 400 mm long, were produced from an automated factory process using a semi-dry concrete mix, thereby allowing efficient forward planning and timely material supply and eliminating all complications and risk that cast in-situ process bears. The minimum crushing strength at 28 days achieved was 30 MPa as per project specifications. The polymeric geogrids used were monolithic, uniaxially orientated HDPE geogrids designed for their 120 years creep strength for the elevated, in soil local temperature requirements as per project specification. The mechanical polymeric connectors were also made of HDPE to provide a high level of load transfer at the grid-block connection at all levels whilst allowing the transfer of horizontal shear loads between adjacent blocks. The shape and feature of these connectors is designed to specifically fit within the modular face blocks and through the geogrids apertures, is durable in all conditions and provides high efficiency connection strength.

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Due to the unprecedented wall heights required in this project, the quality of the reinforced fill material was one of the two most important ‘ingredients’ for the successful completion of this project. The reinforced fill comprised site-won blasted Gabbro, which was crushed down to maximum particle size of 37.5 mm to form a well graded granular fill with minimum Cu of 5. Multiple samples of this material were tested in shear boxes in accordance with the British Standards guidance on soil identification and description. The internal angle of friction was found to be in excess of /’ = 45°. A conservative peak value of /pk’ = 42° was adopted for design. The other ‘ingredient’ that was crucial to the successful construction of these walls was the quality assurance during construction. The actual construction methodology of polymer geogrid reinforced soil walls is not complicated and does not require a specialist; the important part of the construction is to just observing simple rules, like keeping the surface of the concrete modular face blocks properly aligned and religiously brushed clear of any stones, continuity of mechanical connectors and geogrid layers, appropriate tensioning of the often long grid layers (up to 35 m embedment length at the 60 m high tiered walls) and, perhaps most importantly than all, appropriate compaction of the crushed Gabbro reinforced fill. There were many lessons learnt through working in such a vast project, both in wall height and in overall face area. All projects are liable to changes once transferred from the design board to the site but in this project, due to its enormity and variability in the local topography, a lot of tie-in detailing had to be re-assessed and re-designed, once the project became alive on site, in order to save in un-necessary blasting by re-aligning the wall edges, or to accommodate drainage pipes at skew angles near the finished road levels. So one of the lessons learnt was to be extra agile and prepared for extra flexibility and open-mind-attitude when going into such a vast project with challenging topography. But the most important lesson learned, perhaps, was one to do with drainage! In some of the 3-tier walls and in order to meet the require freeway alignment by utilizing the reinforced soil walls to fit in the most efficient way the local topography, the lower tier had to be constructed up to 30 m way from the upper 2nd tier making the 2 structures independent, in design terms, but still requiring a continuous construction. To make the construction as cost and environmentally efficient as possible, we used a lesser quality crushed gabbro fill, i.e. with wider fill particle degradation, to construct the core of the massive earth embankments that were retained by the reinforced soil walls. By doing so, however, we created a potential path for surface water to ingress and as the reinforced fill was of tighter degradation and therefore more closely compacted and therefore more impermeable than the retained backfill, the creation of potential and hard to predict in magnitude water pressures at the back of the reinforced soil block. Once this issue was realized we resolved it by specifying a cement-soil mix layer to be laid at all wall crests exposed to surface rainwater in order to minimize excessive infiltration. The lesson learned was that, in every project, but in particular in large, complex projects, the designer must ‘step back’ and have a holistic view at the whole project and what issues might influence the performance of the overall scheme as a whole rather than just ‘getting lost’ in the detail (Doulala-Rigby and Wills 2011).

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Malaysia Landslide Repair, 2004

On 31st May 2006, a large landslide occurred of the slope that supported a row of twostorey terrace houses at crest located at Taman Zooview in Malaysia. The landslide was reported to have damaged a newly constructed Anchored Soil Wall of 6 m high and 40 m long located at the toe boundary of the development to Taman Zooview, which collapsed and moved about 100 m down the slope causing extensive ground movement and resulting in the destruction of three long houses at the toe of the slope and loss of four lives. Additionally, fifteen (15) units of the 2-storey terrace houses located at the crest of the slope were subject to evacuation order by the local authority as these houses were considered unsafe for occupation. The fifteen (15) units of terrace houses located at the crest of the slope were constructed on loosely placed fill of over 13 m thickness over a ravine leading to frequent not so major land sliding incidents over the last 20 years since the houses were built. The drainage system of the slope was also identified as not functioning either. In addition, local residents developed a ‘habit’ of indiscriminately and illegally discharge domestic and storm water over the slope from their houses. According to various testimonies from the house owners, the fill-slope below their backyard had been ‘eroding and slipping’ since completion and had become critical when the edge of the slope crest came started moving dangerously close to the houses. A slope stability study was undertaken and revealed that failure could be expected to happen at any time, particularly with incessant rainfall and poor drainage system and the factor of safety for localized slope failure was just over 1.00. The unstable slope was left as playground in the proposed new development below the boundary of Zooview houses. When the residents of the terrace houses complained to the local authority on the slope movement, a site meeting was held at which the local authority directed the new developer to stabilize the unstable slope with a 40 m long  6 m high reinforced soil (RS) wall. As the slope was declared already border-line stable with the potential of a landslide imminent, the developer was ordered to design and implement the remediation plan in very short time. The RS wall was built on top of a reinforced concrete slab supported on RC piles at the boundary by a specialist contractor. According to the global stability analysis submitted by the specialist, the wall had a factor of safety greater than 1.4. The design was reviewed and approved by an independent reviewer appointed by the local authority. The remediation scheme did not include any improvement of the problematic drainage condition near the crest of the slope, which remained the same at the time the remedial RS wall + piles were built at the slope toe. The local authority wanted to complete the wall urgently for fear of the raining season, the construction of the wall took about 6 weeks to reach its full height. 10 days after the wall reached its full height of 6 m, on 31st May 2006, the whole slope supporting the backyard of the terrace houses and the new wall + piles at the toe collapsed. The site was waterlogged at the bottom of the landslide while spring water could be seen flowing steadily out of the slope. The landslide material flowed down the slope and caused massive upheaval of the ground below the toe of the slope resulting in the destruction of squatter houses and death of 4 persons. It was at that point the importance of properly designed drainage was recognized. The remedial works were assigned to a reputable company, who founded the new

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remedial works on a substantial subsoil drainage system in the form of a rock-key embankment at the toe of the slope so that the water could be drained away by gravity flow. The rock-key was taken down to firm layer of weathered granite at 64 m AOD, as shown in Fig. 3 below. Subsoil drainage pipes were provided to take the underground water away from the slope toe and into an open drain.

Fig. 3. Typical cross section (after Ooi 2008)

During excavation to found the rock-key onto firm formation, it was revealed that the piles of the previous inadequate remedial scheme did not reach deep enough but they stopped just above or within a layer of soft organic material of about 1 m thickness the was encountered overlaying the firm layer of weathered granite. This layer was believed to have been the weak sliding surface at the toe of the slope that triggered the landslide that caused the upheaval of the ground at the toe. This weak soil layer was subsequently completely removed. The construction was carried out into 5 stages of excavation to make sure that the whole toe was not exposed at once causing further instability. Slope stability was carried out for each excavation stage to ensure safe and controlled construction. Inclinometers on slope and tell-tale glasses on cracks of the existing terrace houses were installed to monitor any slope movement. The finished level of the rock-key that formed the founding level of the reinforced soil slope was at 78.5 m AOD, creating a robust and safely keyed-in, free draining lower rock embankment to support the new reinstated slope above. At this level, i.e. 78.5 m AOD, the reinstated fill embankment was founded and constructed with HDPE uniaxial polymer geogrid reinforcement, that was chosen based on its rigidity, integrity and strength at the junctions of the geogrids as well as its ability to mobilize its strength at compatible low strain levels with the site-won cohesive fill that was recovered from the landslide and was used as reinforced fill. An added lesson learnt, apart from reiterating, once again, the need for proper site investigation to appropriate depths and the holistic and critical consideration of drainage in any earthwork project, was that when using site won fill appropriate rate of construction needs to be allowed for to cater for dissipation of pore water pressure of the compacted cohesive fill, which is vital and must be observed at all cost, when necessary, even if it will result to delayed completion of earthworks. It is better to have a late but robust structure rather than a rushed and unsafe one that is destined to fail,

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as this project sadly demonstrated. The completed slope was 60 m high and reinstated the original size of the backyard of the 15, two-storey housing units at the crest (Ooi and Ting 2005). 3.3

Monserrat Runway Slope, 2006

The original Montserrat W H Bramble airport, located in the Caribbean Sea, was destroyed during the eruption of Soufrière Hills Volcano in 1995. The new airport was planned to be relocated to a safer area in the north of the island. The terrain here was mountainous and extremely undulating and it was necessary to raise the existing ground levels by 30 m to support the new runway. To achieve the required topography, 45o face angle reinforced soil slopes were decided to be constructed at both ends of the runway in order to prevent the need to “chase” the toe of the slope as the natural ground fell away quite steeply, which would necessitate the importation of large quantities of additional fill material. The slopes were designed and constructed utilising HDPE polymer uniaxial geogrids, which provided the most time and cost-effective solution on such a critical project, while facilitating the use of local site won fill material. The construction of the new airport re-named as Gerald’s airport, and terminal comprised a 600 m long airstrip and asphalt surfaced link taxiway, aircraft parking apron, helipad and a corrugated steel tunnel beneath the proposed runway including a new carriageway and pedestrian footpath. In order to provide the 600 m long runway it was necessary to construct two very large reinforced soil slopes; one at the eastern end 12 m high at 450 and the other at the western end some 31.5 m high at 45o as Fig. 4 shows. The fill material used in the structures was as dug Volcanic ash. The scheme was a design and supply project from the geogrid manufacturer and a British consultant. The design paid particular attention to the fact that Montserrat is a tropical island with very heavy seasonal rainfall and is also actively volcanic; therefore, both ground water and seismic loading were a major consideration in the design. To deal with the consequence of high seasonal rainfall the design team used elevated pore water pressure values (ru) within 3 m of the slope face. Designs using Bishop’s Simplified method to assess internal, compound and global stability were considered with a minimum target factor of safety of 1.3 for the static condition. For the dynamic (seismic) condition a target of 1.05 was used for both internal, compound and global factors of safety. The site won volcanic ash was reinforced with HDPE polymer uniaxial geogrids and polymer geogrid surface erosion protection secured at the surface of the finished slope. Both embankments were profiled at 45o with no intermediate berms and planted with native grass species to form an erosion resistant yet attractive finish, which blends well with the natural surroundings of the area. Gerald’s airport was opened in June 2005, providing a vital lifeline to the people of Montserrat and allowing them to look to the future with greater hope and also help rebuild the tourist trade, which was so badly affected after the catastrophic eruption of 1995 (Tensar archives 2005).

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Fig. 4. View of the 12 m and 31.5 m high eastern and western airport embankments

3.4

Edan Valley Working Platform, 2018

The Elan Valley Aqueduct (EVA) was built over a 100 years ago to bring water to Birmingham and surrounding areas from mid-Wales in the UK. The need for modernisation and extensive refurbishment was met with an extensive earth works scheme that included the construction of various access shafts and a new tunnel with a 150 t tunnel boring machine (TBM). To construct a level and horizontal working area to support the construction traffic and a 1,000 t crane, which would be used to assemble the TBM for the construction of the tunnel, a reinforced soil retaining wall (RSRW) with polymeric geogrids and site won material from the shafts was proposed. The length of the RSRW was 160 m, with a maximum height of 13.3 m, a slope angle of 85° and 43 layers of geogrid spaced at 300 mm vertical spacing. A working platform was constructed, on top of the RSRW, with selected granular material of 970 mm thickness and 3 layers of geogrid. An aerial view of the slope under construction is shown on Fig. 5 below.

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Fig. 5. Aerial View of the construction of RSW at Bleddfa

One of the main challenges of the project was to achieve a cut-and-fill balance, so the design of the RSW was based from the beginning on using on-site won material, instead of importing fill material. Based on the results of the preliminary soil investigation, the on-site won material was described as clayey slightly silty sandy gravel and classified as Class 6F1 – fine graded selected granular fill – according to the UK Specification for Highway Works MCHW Series 600. The initial parameters of the fill material used for the design were: characteristic friction angle of 35°, unit weight of 18 kN/m3 and a cohesion of 0. However, according to the test results of the actual on-site available material, the fill was described as dark brown very clayey sandy gravel and slightly gravelly clay and classified as Class 2C – stony cohesive fill – with lower friction angle and larger fines content. The parameters to be used in design where therefore revised to: characteristic friction angle of 28°, unit weight of 18 kN/m3 and cohesion of 0. This material was identified as very susceptible to weather conditions and it was recognised that it would require close placement control and monitoring during its installation and compaction. A detailed and thorough site testing regime was therefore prepared to be undertaken in each compacted reinforced fill layer with regards to minimum compaction requirements, moisture content and Californian Bearing Ratio (CBR) values. The testing regime assigned to be followed during the construction of the RSW using the on-site won ‘dark brown very clayey sandy gravel and slightly gravelly clay’ material comprised soil tests according to BS1377: Part 9:1990 and was as below:

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• Plate Bearing test: minimum four tests per compacted soil layer to obtain the CBR. • Hand Vane Shear test: minimum three tests per layer. • Core Cutter test: minimum three tests per layer. Measure of bulk density and dry density to obtain the relative compaction. Test also used to control the moisture content. • Sand replacement density test: minimum two tests every three days. Test mainly use to compare results of relative compaction and moisture content in laboratory. The minimum requirements for each layer were specified with the following values: • Minimum CBR: 15% • Relative Compaction: minimum 95% of maximum dry density (maximum 5% air voids at a dry density equal to 95% of the maximum dry density from 4.5 kg hammer compaction test) • Maximum moisture content: 12% The plate bearing test, hand shear vane and core cutter were acting as an indicator for the site won performance daily. If the test results were below the minimum requirements, the layer was to be removed completely and replaced with on-site won material from a different batch and retested. The above testing regime was followed for each layer of the RSW. When construction started it was soon realised that the above minimum requirements had to be further refined as the first layer for which results were below the requirements was layer number four already. In this layer the in-situ CBR value were found to be in the order of 10%, with a relative compaction of 85% and moisture content of 18%. The results did not quite meet the minimum requirements; however, they were not so far from the target values (CBR 15% and relative compaction 95%). Rather than excavating and replace the layer as originally planned, and as the weather was dry, the layer was left exposed for two days, then compacted further by some additional compaction passes and then re-tested. The new test results achieved the CBR target of 15%, relative compaction of 95% and the moisture content went down to 12%. So for layers with CBR results between 10% and 15%, relative compaction between 80% and 90% and moisture content up to 18%, and when the weather was dry, the remediation technique was for the layer to be allowed for a couple of days exposure and additional compaction. However, when there was continuous (and torrential) rain for more than three or four days and when the tests results were much lower than the minimum requirements, such as: CBR values less than 10%, moisture content above 25% and relative compaction less than 70%, the procedure followed was to: • Excavate the first 2 m from the face of the RSW down to 300 mm and replace the on-site won material with a lean mix concrete. • Replace all the rest of the layer with crush and run material (Granular material Class 6F1). • Next layer of 300 mm to be installed with on-site won material, with layers compacted each at 150 mm. Additionally, to place a drainage geocomposite in strips of 2 m, in 2 m spacing, between the 150 mm layers.

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The above site-born remedial techniques allowed the subcontractor to work in other sections of the RSW, reducing the delays in the overall construction program. The RSW was successfully completed with a total of 43 compacted soil layers. All layers were tested according to the testing regime described above and following the additional remedial procedures as necessary. On average, the obtained CBR values were between 18% and 20%, with a relative compaction of 97 to 98% and a moisture content between 10% and 13%. The top three layers were part of the working platform and were constructed with imported granular material. The in-situ tested CBR values of the top layers were between 30% and 35%. In December 2016, the TBM arrived at the end section of the tunnel at Bleddfa, after five months of work to complete 1.8 km of tunnel. The 1000 t mobile crane and the heavy loadings were used to extract the TBM from the reception shaft and the platform on top of the RSW performed well, as expected (Guerra-Escobar and Bernardini 2018). 3.5

Greater Bargoed Slopes, 2008

In the heart of Bargoed in Wales, UK, in 2008, a planning permission was granted to build a 22 m high plateau out from a hillside with a 60-degree face angle as part of a community regeneration retail development project providing an additional 12,000 square metres of development land. After careful consideration of all possible alternatives the preferred solution was to offer real cost savings by re-using site won colliery spoil reinforcing them with HDPE polymer geogrids. A rigorous testing regime was undertaken that led to the design parameters being established as c’ = 0 kPa, /’ = 32o and c’ = 19 kN/m3 for the re-use of the available colliery spoil. The roles and responsibilities were clearly defined at the onset of the project with the geogrid manufacturer responsible for the design of the internal stability while the contractors specialist sub-consultant was responsible to tie the manufacturer’s design into the overall scheme including global stability checks of the whole retail plateau. Besides the 22 m reinforced soil slopes 2no associated reinforced soil abutment walls were also designed, which formed the integral bank seat for the north west end of the new Bargoed Newydd Viaduct. The reinforced soil bridge abutments were constructed with granular fill with design soil parameters of c’ = 0 kPa, /’ = 38o and c’ = 19 kN/m3 to avoid any post construction settlement. The 60° slope face was achieved with welded galvanised steel mesh panels connected to HDPE polymer geogrids with rebars though the face. The selection of face options were such to help the slopes blend into the surroundings, with the majority of the side slopes being seeded and vegetated while underneath the bridge structure, for example, where no vegetation could establish, a steel faced stone clad option was adopted. The completed design is amongst the highest tiered reinforced soil embankments ever to be constructed in the UK and the highest to make use of a recycled material (colliery spoil). Over 6,000 m2 of reinforced soil face was installed between 2007/08. As well as providing significant cost savings to the project, this first-class example of sustainable engineering has resulted in major environmental benefits, with approximately 150,000 t of quarried stone being replaced by what was previously classed as a

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waste material. There are also major benefits to both the local and the general environment with approximately 7,500 lorry movements removed from the local road network. Of particular interest in this project was the fact that the colliery spoil was unburnt and therefore potentially susceptible to combustion. In order to overcome this challenge, 1 m wide capping layer of granular fill was installed behind the face thereby protecting the unburnt colliery spoil. A 150 mm wide layer of top soil was placed on top of the granular capping layer and the face of the slopes, facilitating fast vegetation growth that transformed the finished look of the reinforced to an aesthetically pleasing and naturally blending structure into the green surroundings, as shown in Fig. 6 below (Tensar archives 2008).

Fig. 6. Typical view of the complete, 22 m high colliery spoil reinforced soil slope

3.6

Dan-Y-Lan Landfill Landslide Remediation, 2011

Danylan landfill, which is located in Wales, UK, operated from 1955 until its closure in 1971 just prior to introduction of legislation. In early 2004, and following a prolonged period of rain, an approximately 18 m high landslip occurred at the north-western end of the former landfill. The landfill involved approximately 8,000 tonne of uncontrolled refuse tip material in a matrix of made ground. Testing of the landslide debris indicated that, if left unattended, its chemical composition could pose a significant chronic risk to

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human health, flora, fauna and controlled waters. Restoration of the landslide to its original profile included the engineered placement of the failed waste debris back into the slip scar with the aid of polymer geogrid reinforcement providing a robust, cost effective, sustainable, environmentally friendly and maintenance free engineering solution. The location of the slip was on a steep 1:2 (V:H) slope above the River Rhondda, where a mixture of ground water intrusion and standing surface water had destabilised the slope. It was concluded that the landslide was most likely triggered by either raised groundwater levels possibly occurring as a consequence of, or exacerbated by, surface water intrusion via tension cracks close to the slope’s crest. A Desk Study as well as an intrusive geo-environment study were commissioned soon after the landslide to investigate the chemical and physical properties of the waste material. The study for chemical properties concluded that significant pollution linkages to PAH, PCB and isolated areas of lead and nickel contamination were identified. The intrusive soil exploration that was carried out by trial pits, boreholes and laboratory testing through the failed parts of the landfill revealed Made Ground, which comprised predominantly flow debris from the former landfill, overlying Glacial Deposits. The intact layer of the Made Ground at places outside the landslide was found to be up to 10 m thick. The consistency of the Made Ground encountered comprised a heterogeneous mix of fine and coarse-grained soil with variable quantities of waste material including fragments of wood, plastic, glass, fabric, metal, and bone. A car and a fridge were also encountered. The gravel size particles were described as being angular to sub-rounded of mudstone, shale and sandstone origin. A Geotechnical Consultant was appointed to review and interpret the geotechnical field records of intrusive ground investigatory and laboratory works, to provide advice on the selection of effective shear strength parameters for the disturbed landfill materials for use as reinforced fill and the underlying Glacial Deposits for use in limit equilibrium analyses of slope stability and to carry out the external stability of the proposed reinstated reinforced soil structure, thereby providing the required reinstated reinforced soil block geometry to satisfy external stability. As often the case, the polymer geogrid manufacturer was appointed to carry out the internal stability of the proposed reinforced soil structure that was to be reinstated, based on the provided geometry and the soil properties of the disturbed landfill material. The geogrid type chosen was to satisfy internal stability was an HDPE uniaxial geogrid due to its chemical inert nature towards the various contaminants identified in the waste reinforced fill, with approximate long term strength of 24 kN/m, detailed at vertical spacing of 1.0 m with lengths ranging from 2 m to 22 m. Secondary biaxial PP geogrids, 2 m long, were also provided at 0.3 m intervals to assist with the otherwise unsupported face construction. The slope was designed for 120 years design life and the in-soil temperature was taken as 10 oC. The 8,000 tonnes of failed waste material was excavated from where it was initially deposited and was stockpiled on site. The collected waste material went through a process of light screening for removal of large objects. The construction sequence involved levelling of the foundation level and appropriate benching of the back scarp at benches 1.5 m wide. A geocomposite membrane was firstly laid along the base of the slope and back slope benching and up to the maximum level of where groundwater was

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encountered. Free draining granular material was then laid on top of the geocomposite at layers of 150 mm to 300 mm thick and in parallel with the waste fill/geogrid placement. A filter membrane was finally laid on top of the free draining granular material completing the backslope drainage construction that was designed to minimise the inflow of groundwater into the emplaced debris materials, subsequently minimising the leaching potential of contaminated material. A perforated pipe was also installed in a rubble drain constructed along the entire toe of the reinstated slope to collect and safely discharge of the drained water. The bulk of the reinforced slope construction involved placement of the waste fill at 150 mm layers and compaction near to the tested optimum tested density of the fill. Installation of the appropriate type of geogrid at the appropriate level followed and the process was repeated until the slope reached the required crest level. In line with the findings of the geo-environmental risk assessment recommendations, where applicable, debris material was delineated and materials with higher levels of contaminants were placed at depths greater than 2 m from the upper surface of the engineered slope. The engineered embankment was subsequently overlain with a cover system of imported material – thus providing a break in the source-pathway-receptor linkage. The cover system generally comprised 150 mm to 200 mm of topsoil to BS3882:1994 (General Purpose) and 300 mm to 350 mm of subsoil to BS 3882:1994. A warning barrier in the form of a non-woven geotextile was placed beneath the cover system. This was to discourage excavation below the cover system, reduce the degree of intermixing and inhibit root penetration. In order to counteract the downslope migration of the topsoil covering required on the re-instated embankment, rows of brushwood faggot fascines were installed in rows across the embankment. This system subsequently held the topsoil in place until it was hydroseeded with a coir fibre mulch to give an effective protective covering to the slope and minimise erosion. Vegetation of the slope was hence rapid, giving a decent covering at 35 days, as Fig. 7 shows. Following completion of works, sampling and subsequent laboratory analysis of imported materials used in the formation of the cover system was undertaken in order to verify that remediation had been successful. From the test data collected and evaluated it was concluded that the remediation objectives have been successfully met and that no further monitoring was required. Due to the unusually variable and high risk nature of the reinforced fill, as well as the unknown compaction techniques of how to handle such fill, a very close and trusted co-operation between the asset owner, design consultant, constructor and geogrid provider sub-contractor was established from the very beginning of the project, which was the catalyst to the successful implementation and completion of this project. The team was in almost daily contact as a lot of project specific decisions had to be done after the project went on site and based on the observational method. One of the main challenges encountered was the optimum compaction method of the landfill material. The material was originally classified as ‘granular’ but when tried on site it proved challenging to be compacted using traditional granular fill compaction techniques such as plant with vibrating rollers. Due to the nature of the waste fill material the vibrations during compaction lead to moisture concentration at the surface of the compacted layer resulting to dump trucks sinking. To overcome the compaction problem multiple passages of a smaller tracked plant was engaged as an alternative construction method,

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Fig. 7. Vegetation growth on the restored embankment at 35 days.

which added to the overall construction time but was accepted by all parties involved as it effectively solved the problem. Small tractor dumpers also replaced large dumper trucks for transporting the waste within the site (Doulala-Rigby and Stone 2011a, b). 3.7

Tinsley Walls, 2016

In late 2013, a new major road construction development in the North of England, UK was granted full approval by the Department of Transport in the UK. The new road, also known as Tinsley Link, included the construction of a new bridge over River Don and crossing the Supertram tracks, known as the Fitzwilliam Bridge, and the design and construction of approximately 490 m long reinforced earth approach embankments to the bridge on a piled foundation. The two reinforced soil (RS) retaining embankments either side of the bridge were formed with near vertical sides rising to a maximum height of 11.2 m and 7.0 m respectively abutting the bridge abutments. The RS embankment to the northwest of the bridge was the longest, comprising some 460 m long, near vertical, back to back, hard facing RS walls ranging in height from 2.2 m to 11.2 m. The reinforced earth embankment to the southeast of the bridge comprised some 30 m long, near vertical, back to back, hard facing reinforced soil walls ranging in height from 4.8 m to 7.0 m.

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In order to make the design of Fitzwilliam Bridge more cost effective, geogrid reinforced soil thrust relief walls (TR2 walls) were utilized behind both bridge abutments in this project. The TR2 walls were designed to provide lateral thrust relief to the bridge abutments by supporting the horizontal earth pressure from behind the bridge abutments. So in effect, the impact of earth pressure was not considered in the bridge abutment design, making it a lot more cost effective. The existing founding soils comprised low bearing foundation strata that would be unable to support the proposed RS embankments. After consideration of various options, the most cost effective method of ground improvement was deemed to be a combination of Vibro Stone Columns (VSCs) under the lower parts of the wall and concrete piles, under the higher parts of the approach RS embankment with a geosynthetic reinforced granular load transfer platform (LTP) on top, forming the founding medium for the proposed RS approach embankments. The reinforced fill material used to construct the RS walls supporting Tinsley Link was pulverized fuel ash (PFA) reinforced with uniaxial polymeric geogrids. The PFA, which is a waste product of pulverized fuel (typically coal) fired power stations, was supplied by nearby power stations. The choice of PFA as reinforced fill was due its low density, when compared to natural granular or cohesive fill, thereby further minimizing the amount of foundation piling needed. Following bespoke PFA sample testing, the soil properties used for the PFA in the RS wall design were c’ = 5 kPa, /’ = 27o and c’ = 15.5 kN/m3. It is well understood that PFA undergoes a pozzolanic reaction and cements over time, hence the use of an apparent cohesion in design. When newly produced, PFA is strongly alkaline; a pH as high as 11 is known, and >9 is normal but it does often decline in time towards the average value of 7. This high alkalinity can be the reason to exclude some materials from being used to reinforce PFA, such as steel and polyester. Or appropriate reduction factors can be used in design but with extreme care. This problem does not arise with high density polyethylene uniaxial geogrids, which is what was used in this project, as they are inert to chemical attach and are unaffected by high pH conditions. The maximum PFA particle size considered for design was 14 mm. The walls were constructed with modular concrete facing blocks laid dry without using mortar and attached to the geogrids via a polymeric connector similar to the one described in case study 3.1. The grids are laid always with the apertures perpendicular to the wall face blocks and slightly tensioned to remove any slack, as shown in Fig. 8 below. PFA fill was then placed and compacted to 200 mm thickness, typically to 95% of its dry density. The geogrid laid in-between the compacted PFA fill layers at a vertical spacing typically varying from 200 mm near the toe of the wall to 600 mm near the crest.

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Fig. 8. Geogrids being slightly tensioned to remove slack during installation

For Tinsley Link a few extra construction measures had to be implemented in order to facilitate the successful installation of the PFA fill. PFA is susceptible to scour and washout. Additional drainage measures were made as outlined in BS8006-Part 1:2010 Cl. 6.10.5.2, Cl. 6.10.5.3, and Cl. 6.10.2.6.3. The whole PFA reinforced soil block was encased in drainage layers as follows: • A 200 mm thick drainage layer was placed over the entire base of all reinforced soil walls directly above the LTP. • A vertical 300 mm thick drainage sand layer of grading C or M (as defined in BS882) was placed directly behind the modular concrete blocks. • A vertical 1000 mm thick drainage sand layer of grading C or M (as defined in BS882) was placed directly behind the steel mesh facing thrust relief walls behind both the bridge abutments. • A 500 mm well graded granular fill was placed at the crest of the reinforced soil walls in between the PFA and the finished pavement construction. All drainage layers were constructed with adequate drainage pipes that were integrated and discharged off appropriately to the wider site drainage system away from the embankments. Movement joints 20 mm wide filled with compressible silicon joint sealant were constructed along the embankment where the foundation type changed from VSCs to PRIs and at the interface between the edge of the bridge abutment pile cap and the abutted remaining MSE walls. One particularity of this scheme, and a good practice for other future schemes, was the use of different reinforced fills to accommodate the project particularities. As the wall was located on a flood plain, and in order to avoid PFA leaching if submerged, the lower part of the walls and up to the given flood levels was constructed with well graded granular fill (Doulala-Rigby and Black 2015).

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FARRRS Walls, 2017

In June 2012, a new strategic highway project in England was granted planning permission. The new highway, known as Finningley And Rossington Regeneration Route Scheme (FARRRS), is a 3 mile highway that was to provide a strategically important connection between junction 3 of M18 motorway to Doncaster Sheffield Airport, UK. The scheme included a group of 6No. mechanically stabilised earth (MSE) retaining wall type structures forming 2No. approach embankments to a new bridge to carry the new FARRRS highway over the existing East Coast Main Rail Line (ECMRL). The 2No. approach embankments either side of the bridge were formed with near vertical sides (i.e. 4o slope from vertical) MSE back-to-back modular concrete facing block walls rising to a maximum height of 12.8 m and 12.5 m respectively and abutting the west and east bridge abutments. The existing founding soils comprised variable, low bearing capacity foundation strata that would be unable to support the proposed high reinforced soil embankments and would potentially be prone to differential settlement. The loading of the proposed new embankments, if constructed from conventional heavy fill, would potentially also have detrimental effects on the stability of the existing ECMRL tracks. To mitigate the settlement effect on the ECML tracks from the embankment loading and in order to reduce the remediation effort for the existing foundation soils, it was proposed to use a reinforced soil solution using LWA fill material. A number of alternatives were previously considered for the construction of the approach embankments. One of them was the use of polystyrene fill but was later dismissed as economically unviable for this scheme. Additionally, a 3-span increasing height bridge that would have replaced the need of the 2No. approach embankments over the weak foundation soils was also considered but was found to be disadvantageous for this scheme in terms of cost and programme and future maintenance liabilities. The chosen embankment fill material was Leca® LWA (LLWA), which is an expanded, lightweight clay granular ceramic material. The clay is pelletized, dried and expanded in rotary kilns. The output LLWA granules, in the range 0–32 mm in diameter, are sieved into different products. Like any other granular material, LLWA offers good frictional resistance without any cohesion, with an effective angle of internal friction of /’ = 36o and bulk density of c’ = 5 kN/m3. External stability checks comprise sliding, bearing and overturning/middle third checks. Due to the low density of the LWA, the reinforcement lengths required to achieve target factor of safety for sliding are higher than the routine length of 70% of the mechanical height of the structure that is normally expected when naturally sourced granular fill material is used. Due to the low bulk density of the LWA and hence low weight of overburden, bearing capacity factors of safety were easily met and the middle third rule was equally satisfied. Due to the low weight of LLWA used throughout the approach embankment footprint, settlement and differential settlement was estimated to be minimal and within the allowable limits, as expected. Internal stability comprises two fundamental checks: failure against rupture and failure against pull-out of geogrid. Internal stability provides the geogrid strengths and spacing that are required for a stable structure. The interaction factors used for sliding

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and pullout cannot be assumed as they are specific to proprietary geogrids and LLWA products, that had already been tested together in the past, and must be only derived by product-specific testing. The use of previous testing of the interaction factors of the specific proprietary geogrids and the LLWA fill is perfectly acceptable as both the geogrid and the fill material are man-made and produced in a quality-controlled manner ensuring uniformity of both products and hence test results repeatability. Overall, due to the lower density of the LLWA fill, the critical internal failure mechanism is sliding rather than rupture, as expected. The construction of a polymeric geogrid reinforced retaining modular block wall is a relatively straightforward procedure with all components dry laid, in accordance to a well-established and tried-and-tested construction sequence. However, when LLWA is used, specialist advice from the LLWA provider must be sought with regards to handling, placement and compaction. LLWA was placed by tipping and spreading using a tracked dozer or tracked excavator and a spreading plant of a maximum ground bearing pressure not greater than 50 kN/m2 for the 10/20 mm grade LLWA used. Compaction of LLWA material was conducted using a tracked vehicle with a maximum ground bearing pressure of 50 kN/m2. Typically, 3–4 passes of such tracked plant were required to achieve adequate compaction. Such compaction typically provides a reduction in volume of 8–12% with the average being approximately 10%.

Fig. 9. Panoramic view of complete FARRRA MSE Walls.

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The vertical geogrid spacing in this project was every 3 facing blocks (450 mm) and the maximum embedment length was in the order of 11 m. The whole scheme was completed in mid-2016 and the MSE walls can be seen operational in Fig. 9 above (Doulala-Rigby et al. 2017) 3.9

EPS Embankment, 1990s

There is no ‘official’ design guidance through any national or international code or standard on the use of EPS as fill reinforced with geogrids, although the combination of the two has been used successfully in a number of projects, such as the London Olympics in 2012. As often is the case, the used of EPS is likely to be an ‘after thought’, perhaps when loadings were underestimated so EPS is utilized to minimize the overburden loading over undersigned or unprecedented weak foundations. It is a good solution but needs to be thoroughly considered and the design needs to be bespoke to the specific project. The case study that is presented here briefly summarises the combined use of EPS and geogrids. The use of geogrids here were not to reinforce the EPS blocks but to create a stable ‘seating platform’ for the EPS slope core to be founded upon and to mitigate differential settlement, which is another solution where geogrids can work in synergy with EPS block light with fill.

Fig. 10. Typical section through Irwell replacement bridge with EPS embankment.

The original bridge on the Manchester to Liverpool line over the former alignment of the river Irwell, which was diverted into the Manchester Ship Canal in 1900, was constructed in 1892. The bridge had reached the end of its service life and train speeds had been downgraded to 25 mph so replacement was needed. The old riverbed was infilled with up to 8 m of very soft clays and silts and was contaminated with arsenic, hydrocarbons and methane so removal was too difficult, but any solution needed to be

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light enough to avoid excessive consolidation. The solution chosen was the use of EPS blocks to form an embankment to support the rail line and replace the existing bridge all together. A geogrid stabilised granular basal layer and methane cap was placed directly on top of the infilled riverbed before a 14.5 m high EPS embankment was built up to the level of the bridge soffit and 1:1 granular soil embankment sides placed, as Fig. 10 shows. The bridge was then removed prior to placement of the final layers, ballast and track reinstatement and the rail line reopened at full network speeds (Woods, personal communication 2018). 3.10

HS1 Rail Embankments (Previously Known as CTRL), 2003

Back in 2003, the existing North Kent railway line (NKL) had to be connected to the new high-speed Channel Tunnel Rail Link, or CTRL as known back then, by constructing an 8 m high embankment over soft, environmentally sensitive marshland. Polyester high strength geotextile reinforced load transfer platform 1 m thick, spanning driven, cast in-situ piles was used to create the foundation to support the 8 m high, 60° steep, lime-stabilised chalk embankment reinforced with polymer uniaxial geogrids above the load distribution platform, using a wraparound vegetated face, as shown is Fig. 9. The CTRL was Britain’s first true high-speed railway line at the time. It was fully completed in 2007 creating 180 km of high speed rail line linking the Channel Tunnel to London. A new railway embankment, approximately 220 m long was required to connect the existing NKL railway to the CTRL. Locally the site was a marshland area covered with reed beds with groundwater encountered at ground level. Ground conditions comprised up to 1.6 m of weak and variable made ground overlying low strength alluvial clays, to a depth up to 6 or 7 m below ground level. Medium dense to dense gravel was encountered towards the base of the alluvium for the most part, which was underlain by the Upper Chalk Formation at approximately 6 m to 10 m below ground level. The poor ground conditions and tight tolerance requirements dictated the need for a piled embankment. For economic reasons a geosynthetic reinforced platform was designed to spread the vertical embankment loads on to the piles. Driven cast in-situ concrete piles were used and the platform was designed to BS 8006 with a 120-year design life, using a high strength knitted polyester geotextile to reinforce the load transfer platform. In order to limit encroachment onto the ecologically sensitive marshland and also minimise the number of piles, the embankments footprint was reduced by utilizing reinforced soil technology to construct the railway embankments standing at 60-degree from the horizontal. The original design required approximately 37,000 m3 of imported, well graded, granular fill (UK Highways Class 6I/6 J). However, there was a surplus of chalk available on site, albeit with a high moisture content (m.c. > 28%) and low density. A value engineering exercise was therefore carried out, which included trials to investigate stabilisation of the chalk by site rotavation with quicklime to absorb the excess moisture. It was found that 2% lime achieved a m.c. of 23–25% and that compaction to a maximum 10% air voids was satisfactory to meet the design parameters. Cost savings of approximately £700k resulted in addition to the obvious environmental benefits.

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Fig. 11. Typical photo showing the construction of the 8 m high polymeric reinforced slope.

A lesson learnt in this project is the need for a holistic and innovative engineering attitude when planning material use for a reinforced soil project, always opting for the re-use of locally sourced materials, even if they first seem unsuitable. The critical consideration in this project was that the available site-won chalk was overly saturated and therefore non suitable for re-use, unless treated. The flexible nature and chemically inert composition of the geogrid solution adopted made the use of this otherwise unsuitable, site-won chalk possible. The chalk was stabilized with lime creating an exceptionally alkaline local environment, with pH measurements of about 12.5, which was the main factor in deciding the type of geogrids to be used. The geogrids chosen were manufactured from high density polyethylene (HDPE) and could therefore resist extremely aggressive chemical environments and hence were deemed suitable for use with the highly alkaline, lime stabilized chalk. The geogrids were installed to form a 60-degree wraparound face, as Fig. 11 below shows, which retained a hessian lining and seeded topsoil layer, which gave a nice green finish to the railway embankments creating an environmental and aesthetically pleasing solution (Doulala-Rigby and Dixon 2011).

4 Conclusions As with most engineering advancements, the application of a new technology comes first and understanding the exact mechanism of the technology comes after. The use of polymer geogrid reinforced soil technology in geotechnical engineering is no exception. While the successful use of the first polymer geogrid in reinforcing the soil walls in Newmarket/Silkstone walls in the UK in the early 1980’s was based on sound

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engineering judgement and inquisitive engineering spirit, the true understanding of its actual engineering mechanism in the soil came more than a decade later, with the first publication of the ‘official’ design codes following soon after. Since the introduction of these new products back then, that were recently recognised as ‘the innovation that revolutionized civil engineering’ (ICE200 2018), the use of polymer geogrids with good quality granular material has gained wide acceptance over the years forming a reliable and popular soil retaining solution. However, the use of lesser quality or ‘nonstandard’ fills, such as lower quality site-won cohesive fills, demolition waste, coal/steel industry by-products or otherwise waste and other unusual fills, with polymeric geogrid has stifled and has been vastly neglected. This poor statistic needs to change, especially as most developed nations around the world have pledged to work towards more sustainable construction materials and practices, and to ‘Adapt or Die’ the 17 Sustainable Devolvement Goals (United Nations Summit 2015), which form the United Nation’s blueprint to achieve a better and more sustainable future for all by 2030. The author hopes that this paper, through showcasing real projects that have been constructed with alternative, more sustainable reinforced fill materials, will help raise awareness and improve the current poor statistic. The paper also aims to highlight that, unlike other forms of soil reinforcement, like steel strips or uncoated polyester strips that can only be used with unsustainable high quality granular fills or chemically neutral fills respectively, HDPE polymer geogrids can be used to reinforce a wide range of marginal, highly alkaline or acidic soils, or indeed waste fill materials as long as the right controls are in place and the correct construction sequence is followed. By writing this paper the author also hopes to inspire and encourage the future generation of engineers who will design with geosynthetics to stretch the boundaries, take calculated risks and become more innovative with their value engineering proposals by opting for non-standard, challenging but more sustainable fills that can be engineered, as viable alternatives to imported virgin granular fill. Acknowledgements. The author wishes to thank Tensar International Limited for allowing her access to the use of the company’s archives and the permission to using some previously un-seen photos. Special gratitude is expressed to Emeritus Prof. Colin Jones for all his input, peer review and advice towards this paper. The author also wishes to thank Patricia Guerra-Escobar of Geosynthetics Limited and David Woods for providing some of the case studies’ information featured in this paper. Lastly, but equally gratefully, the author wishes to thank the UK Corps of Royal Engineers for their generous assistance in enabling her obtaining permission and escorting her while visiting some of the oldest historic HDPE polymer geogrid reinforced soil blast bunds located within various Army Barracks, that can only be featured anonymously in this paper’s oral presentation.

References Bagir, T.: Iraq Journal, British Museum (1944) Bassett, N.: Prefabrication Roman style, New Civil Engineer, August 1981 Berg, R.R.: Guidelines for Design, Specification, and Contracting of Geosynthetic Mechanically Stabilized Earth Slopes on Firm Foundations, FHWA-SA-93-025 (1993)

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BS 8006:1995 Code of Practice for Strengthened/reinforced soils and fills, BSi London (1995) Coyne, M.A.: French Patent Specification No. 656, 692 (1929) Department of Transport, Reinforced Earth Retaining Walls and Bridge Abutments for Embankments, Tech Memo BE 3/78 (1978) Doulala-Rigby, C., Black, M.: The design and construction of a bridge approach embankment utilising mechanically stabilised earth walls with geogrid reinforced pulverised fuel ash fill. In: Proceedings of 3rd Pan-American Conference on Geosynthetics, Miami, USA (2015) Doulala-Rigby, C., Dixon, J.: Use of site won chalk for the construction of steep geogrid reinforced soil embankments in the South of England, UK. In: Proceedings of 15th European Conference on Soil Mechanics and Geotechnical Engineering, Athens, Greece, September 2011 Doulala-Rigby, C., Karri, S., Branford, R.: The use of polymeric geogrids with light weight aggregate fill. In: Proceedings of 19th International Conference on Soil Mechanics and Geotechnical Engineering, Seoul, South Korea (2017) Doulala-Rigby, C., Stone, A.: Landfill slip failure repair with geogrids using waste fill material at Danylan, Wales, UK. In: Proceedings of 14th Asian Regional Conference on Soil Mechanics and Geotechnical Engineering, Hong Kong, China, May 2011 Doulala-Rigby, C., Stone, A.: Landfill slip failure repair with geogrids using waste fill material at Danylan, Wales, UK. In: Proceedings of 2nd World Landslide Forum, Rome, Italy, October 2011 Doulala-Rigby, C., Wills, P.: Reinforced soil retaining wall systems reach new heights in the middle east. In: Presented and Published in the 15th African Regional Conference on Soil Mechanics and Geotechnical Engineering - Maputo, Africa, July 2011 Ford, H.: Frank Brian Mercer O.B.E., Biographical Memoirs of Fellows of the Royal Society, vol. 46 (2000) Forsyth, R.A.: Alternative earth reinforcements. In: ASCE Symposium, Earth Reinforcement, Pittsburgh (1978) Guerra-Escobar, P., Bernardini, P.: Construction of a reinforced soil wall for a working platform for 1000t crane and TBM on Elan Valley Aqueduct, Ground Engineering Magazine, UK (2018) Institution of Civil Engineers, Proceedings of Conference on Polymer Grid Reinforcement, Thomas Telford, London, UK (1984) Institution of Civil Engineers, Proceedings of Jubilee Symposium on Polymeric Geogrid Reinforcement’a, Thomas Telford, London, UK (2008) Institution of Civil Engineers, Shaping the World: Two Hundred Years of the Institution of Civil Engineers, Tensar Geogrids, ICE200 (2018) Jones, C.J.F.P.: Earth Reinforcement and Soil Structures. Butterworths Advanced Series in Geotechnical Engineering, London (1985) Jones, C.J.F.P., Doulala-Rigby, C.: The first polymeric geogrid reinforced soils structure. In: Published in the 10th International Conference on Geosynthetics, Berlin, Germany (2014) Jones, C.J.F.P.: Personal Communication (2019) Koerner, R.M.: Designing With Geosynthetics, 5th edn. Pearson Prentice Hall, Upper Saddle River (2005) Lallemand, M.F.: French Patent Specification No. 1173383 (1959) Munster, A.: United States Patent Specification No 1762343 (1930) Ooi, T.A., Ting, W.H.: Report on some major geotechnical disasters in Malaysia. In: Proceedings of International Conference Geotechnical Engineering for Disaster Mitigation and Rehabilitation. World Scientific Publishing Company, Singapore (2005) Pasley, C.W.: Experiments on Revetments, vol. 2. Murray, London (1822)

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Author Index

A Amhadi, Talal, 51 B Bathurst, Richard J., 22 Bhattacharjee, Arup, 95

L Lai, Belinda, 1 O Onukwugha, Eze, 51 Onyelowe, Kennedy Chibuzor, 51

C Chen, Xin, 82 Chowdhury, Swaraj, 9

P Patra, NiharRanjan, 9

D Doulala-Rigby, Chaido, 121

Q Qiu, Haojie, 82

E Ezugwu, Charles, 51 G Gómez, J. M. Muñoz, 62 H Hao, Guiyu, 82 He, Hua, 82 Huang, Peter, 1 I Ikpa, Chidozie, 51 Iro, Uzoma, 51 J Jeon, Han-Yong, 108 Jideofor, Ifeoma, 51

S Saikia, Sudipta Sikha, 95 Scuero, Alberto, 68 U Ugorji, Benjamin, 51 Ugwuanyi, Henry, 51 V Vaschetti, Gabriella, 68 W Wilkes, John, 68 Y Yao, Jialiang, 82

© Springer Nature Switzerland AG 2020 F. Tatsouka et al. (Eds.): GeoMEast 2019, SUCI, p. 151, 2020. https://doi.org/10.1007/978-3-030-34242-5